University of South Florida

Scholar Commons @USFTheses and Dissertations

6-1-2006

The use of mini-pile anchors to resist uplift forces in lightweight structuresJulio AguilarUniversity of South Florida

Scholar Commons CitationAguilar, Julio, "The use of mini-pile anchors to resist uplift forces in lightweight structures" (2006). Theses and Dissertations. Paper 2433. http://scholarcommons.usf.edu/etd/2433

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The Use of Mini-Pile Anchors to Resist Uplift Forces in Lightweight Structures

by

Julio Aguilar

A thesis submitted in partial fulfillment of the requirements for the degree of Master of Science in Civil Engineering Department of Civil and Environmental Engineering College of Engineering University of South Florida

Major Professor: A. G. Mullins, Ph.D. Rajan Sen, Ph.D. Abla Zayed, Ph.D.

Date of Approval: November 6, 2006

Keywords: hurricane, wind, tension, soil strength, foundation design Copyright 2006, Julio Aguilar

Acknowledgments I would like to thank Structural Engineering and Inspections Inc. (SEI) for their assistance in this project and Mr. Steve Covey, for permitting the documentation which was carried out. I would also like to thank Dr. Gray Mullins for allowing me to join his research team, without which I would not have been able to accomplish all I have done. I would like to thank all of my friends who have been there when I needed an extra hand, particularly Mr. Daniel Winters, Mr. Michael Stokes, Mr. Newton Casey, Mr. Anthony Vieira, Mr. Joseph Gadah, and Mr. Andrew Schrader. Finally, I would like to thank my parents, David and Carolice Aguilar, and my brother David Aguilar Jr. They have always supported me and always encouraged me to pursue my education as far as possible.

Table of Contents List of Tables ................................................................................................................... iii List of Figures .................................................................................................................... v Abstract ........................................................................................................................... vii Chapter 1 Introduction ....................................................................................................... 1 1.1 Overview .......................................................................................................... 1 1.2 Scope of Project ............................................................................................... 3 1.3 Organization of the Report ............................................................................... 3 Chapter 2 Background ....................................................................................................... 5 2.1 Foundation Loads in Structures ....................................................................... 5 2.2 Determination of Wind Loads .......................................................................... 7 2.3 Different Types of Mini-Piles .......................................................................... 9 Chapter 3 Alternative Foundations .................................................................................. 17 3.1 Determination of Forces on a 60x100x22ft-7in Building .............................. 17 3.1.1 Analysis Using the Simplified Procedure ....................................... 17 3.1.2 Analysis Using the Analytical Method ........................................... 22 3.1.3 Determination of Dead Load ........................................................... 25 3.1.4 Determination of Uplift Force to be Resisted. ................................ 26 3.2 Incorporation of Both Tension and Compression Forces in Footing Design . 27 3.2.1 Design of Bulk Footing to Resist Uplift Forces .............................. 27 3.2.2 Design of Mini-Piles ....................................................................... 28 3.2.2.1 CPT Method ..................................................................... 28 3.2.2.2 Titan Method ................................................................... 29 3.2.3 Compression in the Footing ............................................................ 30 3.3 How Testing Can Aid in Safety and Economy .............................................. 31 3.3.1 Safety .............................................................................................. 31 3.3.2 Economy ......................................................................................... 32 Chapter 4 Construction and Testing ................................................................................ 35 4.1 Site Investigation ............................................................................................ 35 4.2 Field Test ....................................................................................................... 35 4.3 Site Survey ..................................................................................................... 36 4.4 Mini-Pile Installation ..................................................................................... 37 i

Chapter 5 Economy of Foundations ................................................................................. 46 5.1 Mass Concrete Footer .................................................................................... 46 5.2 Mini-Pile Anchor .......................................................................................... 47 5.3 Break Even Analysis ...................................................................................... 49 Chapter 6 Conclusion and Summary ............................................................................... 54 References ........................................................................................................................ 56 Appendices ....................................................................................................................... 57 Appendix A Mini-Pile Location and Structural Plans of Case Study .................. 58 Appendix B Results of CPT Testing .................................................................... 62 Appendix C Capacity of Mini-Pile At Each Location Using CPT Method ......... 75 Appendix D Mix Proportions for Drilling and Casting ....................................... 85 Appendix E Break Even Analysis ........................................................................ 86 Appendix F Capacity of a Mini-Pile Based on Soil ............................................. 92 Appendix G Cost Per Kip Resisted Based on Soil .............................................. 93

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List of Tables Table 2.1 Possible Load Combinations ............................................................................. 6 Table 2.2 Design Pressures for Different Zones and Speeds ........................................... 13 Table 3.1 Adjustment Factor for Building Height and Exposure, 8 ................................ 21 Table 3.2 Simplified Method ........................................................................................... 22 Table 3.3 Analytical Method ........................................................................................... 24 Table 3.4 Dead Load Calculation .................................................................................... 26 Table 3.5 Minimum Pile Length Using The CPT Method .............................................. 29 Table 3.6 Minimum Pile Length for NE Corner Using The Titan Method ..................... 30 Table 5.1 Construction Cost by Item ............................................................................... 46 Table 5.2 Cost for Each 17kip Mass Concrete Footer ..................................................... 47 Table 5.3 Total Cost Using Mass Concrete Footer .......................................................... 47 Table 5.4 Cost for Each 17kip Mini-Pile ......................................................................... 48 Table 5.5 Total Cost Using Mini-Piles ............................................................................ 48 Table 5.6 Cost per Kip Resisted ...................................................................................... 49 Table B.1 CPT Sounding for NE Corner ......................................................................... 66 Table B.2 CPT Sounding for NW Corner ........................................................................ 68 Table B.3 CPT Sounding for SE Corner .......................................................................... 70 Table B.4 CPT Sounding for SW Corner ........................................................................ 73 Table C.1 Capacity of a Mini-Pile in the NE Corner ....................................................... 75 Table C.2 Capacity of a Mini-Pile in the NW Corner ..................................................... 78

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Table C.3 Capacity of a Mini-Pile in the SE Corner ....................................................... 80 Table C.4 Capacity of a Mini-Pile in the SW Corner ...................................................... 83 Table D.1 Mix Design for Mini-Piles .............................................................................. 85 Table E.1 Break Even Analysis of Project ....................................................................... 86 Table F.1 Capacity of a Mini-Pile in Soft Sand ............................................................... 92 Table G.1 Cost Per Kip Resisted (General Soil Properties) ............................................ 93

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List of Figures Figure 1.1 Change in Cement Prices .................................................................................. 1 Figure 1.2 Florida Wind Speed Map .................................................................................. 2 Figure 2.1 MWFRS Wind Influence Zones ..................................................................... 12 Figure 2.2 Classification Based on Method of Construction ........................................... 15 Figure 2.3 Classification Based on Method of Grouting ................................................. 15 Figure 2.4 Relative Relationship Between Mini-Pile Application, Design Concept and Construction Type .......................................................................................... 16 Figure 3.1 USF CPT Truck .............................................................................................. 33 Figure 3.2 Results from the NE CPT Sounding ............................................................... 33 Figure 3.3 Soil Classification Chart ................................................................................. 34 Figure 3.4 Soil Classification for NE Corner ................................................................... 34 Figure 4.1 Installed Mini-Pile .......................................................................................... 40 Figure 4.2 Williams Form Bar ......................................................................................... 40 Figure 4.3 Anchor Installation ......................................................................................... 41 Figure 4.4 Static Load Test Setup .................................................................................... 42 Figure 4.5 Load Test Results ........................................................................................... 42 Figure 4.6 Site Layout ...................................................................................................... 43 Figure 4.7 SPT Drill Rig .................................................................................................. 44 Figure 4.8 Centralization Tabs ......................................................................................... 45 Figure 4.9 Load Transfer Connection................................................................................45

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Figure 5.1 Soil/Site Specific Break Even Analysis .......................................................... 52 Figure 5.2 Mini-Pile Anchor Capacity Based on Soil Type ............................................ 52 Figure 5.3 Foundation Cost Per Force Resisted (Using Titan Method) ........................... 53 Figure A.1 Site Layout ..................................................................................................... 58 Figure A.2 Roof Framing Plan ......................................................................................... 59 Figure A.3 Front and Rear Elevations ............................................................................. 60 Figure A.4 Side Elevations .............................................................................................. 61 Figure B.1 CPT Sounding for NE Corner ........................................................................ 62 Figure B.2 CPT Sounding for NW Corner ...................................................................... 62 Figure B.4 CPT Sounding for SW Corner ....................................................................... 63 Figure B.3 CPT Sounding for SE Corner ........................................................................ 63 Figure B.5 Soil Classification for NE Corner .................................................................. 64 Figure B.6 Soil Classification for NW Corner ................................................................. 64 Figure B.7 Soil Classification for SE Corner ................................................................... 65 Figure B.8 Soil Classification for SW Corner ................................................................. 65

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The Use of Mini-Pile Anchors to Resist Uplift Forces in Lightweight Structures Julio Aguilar ABSTRACT In the state of Florida one of the primary factors which influences design of structures is the effect of hurricane force winds on structures. These forces can be greater than any other force encountered throughout the lifetime of said structure. For this reason, designing a structure to resist such forces can greatly increase the cost and time required for completing construction projects. Traditionally, large concrete footings have been utilized to resist wind-induced uplift forces. These footings do little more than act as large reaction masses to weigh down the building. An alternative and little-used method for resisting these large uplift forces is the use of mini-pile anchors. Mini-pile anchors generate side shear at the interface between the pile and the soil which resists the uplift forces. This thesis provides an overview of the design methods used to estimate windinduced uplift forces and several foundation options used to withstand these forces. More traditional/less complicated foundations are compared to the more sophisticated mini-pile method which makes more efficient use of construction materials. The cost efficiency of each method is evaluated which provides a guideline for where and when a given foundation option is appropriate. Finally, a case study where the new method was used is presented which documents the design and construction procedures.

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Chapter 1 Introduction 1.1 Overview This thesis explores a more cost effective foundation design for resisting the uplift forces generated by hurricane force winds. This as well as other alternate methods are being considered throughout the state because of the increase in the cost of labor and construction materials in recent years, Figure 1.1 is the change in the cost of cement from 1900 to 2002.

Figure 1.1 Change in Cement Prices (Adapted from USGS Mineral Cost of Cement [4]) Additionally, in the wake of the recent hurricane seasons, there have been increased/heightened restrictions on the wind loads which are being used in the design of buildings in hurricane prone areas. The American Society of Civil Engineers Code for

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Wind Loads (ASCE 2002) [1] requires the state of Florida to design structures to withstand winds of no less than 90 miles per hour along the northern portion of the state, and these values increase as the location of the structures nears the coast. For instance, in the southern tip of Florida, the code specifies that buildings withstand minimum hurricane force winds of no less than 150 miles per hour. See Figure 1.2

Figure 1.2 Florida Wind Speed Map (Adapted from ASCE 7-02) These contours show higher design wind speeds than previous codes; making new, more efficient construction methods essential. This phenomenon is particularly problematic for light-weight steel structures where the self weight is not sufficient to offer considerable uplift resistance.

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1.2 Scope of Project The aim of this thesis was to compare the use of mini-piles with that of large concrete footings for their resistance to uplift forces. Also considered was the design method which is used, along with the benefits of these methods. The CPT[2] and Titan[3] methods were both considered to determine if there were any specific benefits which one method may generate. A case study was performed in Bradenton, Florida at a location where an existing structure was being relocated and renovated to meet the newer building codes. This structure was a 60ft x 100ft steel building with a peak roof height of 22ft-7 in consisting of a structural steel internal frame. Soil characteristics were determined by performing CPT tests in the approximate locations of the 4 corners of the structure. A test mini-pile was constructed based on the calculated values to confirm if these values were in line with the field conditions. Finally, production installation of all the mini-pile anchors was performed. 1.3 Organization of the Report This report is organized into five subsequent chapters. Chapter 2 gives the background on applications where tension loads develop in structures and foundations and when they do not. Also included are examples of the different types of mini-piles, and an explanation of how wind loads are determined. Chapter 3 discusses the design of foundations for structures resisting both uplift and compressive forces for the different foundation methods as well as how testing can increase the quality assurance and economy of a structure. Chapter 4 gives a general overview of the structure for which the case study was done. The construction and testing of the mini piles is explained as

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well as the choice of anchor length and steel which was used. Chapter 5 is an explanation of the economy of the alternative foundations. Finally, chapter 6 gives the conclusions which were determined after all of the testing was done and after the structure was completely installed.

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Chapter 2 Background 2.1 Foundation Loads in Structures The loads a foundation is likely to experience stem from a number of sources and manifest themselves in axial compression, axial tension, lateral/shear and/or bending moments. Further, depending on the probability of one or more load types being applied at one time, most contemporary design codes group load types in a variety of occurrences called load cases. Simply stated, load cases assemble all possible load combinations and discard improbable conditions such as people standing on a roof during a hurricane. Typical load types for Florida structures include: permanent structure weight, called dead loads; movable loads like people, furnishings, or equipment, live loads; windinduced loads, both pressure and suction; and water or rain loads. By combining these loads in common combinations a range of possible loadings are developed for a given foundation based on the magnitude of the load and geometry used to withstand these loads. Table 2.1 shows load combinations/cases recommended by the American Institute of Steel Construction (AISC, 2002).

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Table 2.1 Possible Load CombinationsCase 1 2 POSSIBLE LOAD COMBINATIONS Load Com bination 1.4*(Dead Load) 1.2*(Dead Load)+1.6*(Live Load) + 0.5*[(Roof Live Load) or (Snow Load) or Rain Load)] 1.2*(Dead Load)+1.6*[(Roof Live Load) or (Snow Load) or (Rain Load)] 3 + [0.5*(Live Load) or 0.8*(W ind Load)] 1.2*(Dead Load)+1.6*(W ind Load)+0.5(Live Load) + 5 6 0.5*[(Roof Live Load) or (Snow Load) or (Rain Load)] 1.2*(Dead Load)+/-1.0*(Earthquake Load)+0.5*(Live Load)+0.2*(Snow Load) 0.9*(Dead Load)+/-[1.6*(W ind Load) or 1.0(Earthquake Load)]

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In table 2.1 load multipliers (eg. 1.2, 1.6, 0.5, etc.) have been established statistically based on the probably of more than one loading condition occurring at the same instance in time. It is conceivable, because of its size and weight, to overlook the fact that there may be instances throughout the service life where tensile forces may develop within the foundation. In most structures, only wind loads cause uplift loads in a foundation by overturning, pure suction uplift, or a combination of both. The Florida building codes all require that structures be able to withstand pressures from winds for speeds ranging from 90mph to 150mph depending on the location. Figure 1.1 shows the Florida Wind Speed Map wherein the southernmost tip of Florida is most likely to experience the highest wind speeds. These speeds generate wind pressures from 12.8psf to 49.4psf for 90 to 150mph, respectively on the windward side. Also, the velocity of the wind as it passes around the structure can create a vacuum on its leeward side, and these forces can range from -16.9psf to -60.0psf again respectively, as shown in Table 2.2.

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2.2 Determination of Wind Loads Loads such as live load, dead load, and rain loads are relatively straightforward computations and are either simple calculations of volume and density or prescribed live load values based on the application. Wind load computations are more rigorous involving the wind speed, wind direction, surrounding structures, topography, and structural shape/geometry. The first factor is the wind velocity. The direct velocity with which the wind impacts a structure will tend to generate positive pressures on the windward side and negative pressures on the leeward side of a structure. There is a direct relationship between the wind velocity and the wind load, as an increase in velocity will generate a corresponding increase in pressure. The direction from which the wind is impacting the structure will also play a significant factor on the wind loads being analyzed. If the wind is blowing parallel to the shorter walls of a building, the forces generated will be less than that of the larger walls, as there is less surface area. This is not to say that the pressure will be different, as the pressure is a function of the velocity, however the total force which will need to be resisted will be smaller due to the smaller area being affected. The exposure of the building to these forces will also determine the pressures which will be exerted. If the building is located in an area surrounded by trees, if the ground is uneven, or if the building is located on the leeward side of another building being affected, then the wind pressures generated will not be similar to those of a structure which was constructed in a flat open field. This is because the turbulence

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generated by the interaction of the wind with these other terrain features may decrease the pressure which the wind will exert on the structure. The topography of the terrain is also very important when determining the wind loads which will be generated. If a structure is located on the windward side of a hill in an otherwise flat area, then the velocity of the wind as it crosses over the mountain will be greater than that of wind which has been unimpeded similar to how the wind above the wing of an airplane is traveling at a faster velocity than that below it due to the shape. Alternately, a structure located on the leeward side of a hill may have a significant portion of the wind being blocked by the hill, and therefore the pressure exerted across its surface will be less than expected based on the wind velocity. The rigidity of a structure also plays a key role in the force which will be exerted during wind gusts. The dynamic impact of gusts on rigid structures is less significant than that of flexible structures. This is because the gusting will be more likely to generate movement in a flexible structure than in a rigid one, and this movement can lead to a failure of the system to resist the loads being exerted upon it. When these factors are taken into consideration along with others, an accurate picture of the interaction of the wind with a structure can be determined. The ASCE 07 standard for analysis of wind loads takes all of the above properties into consideration along with the Importance Factor, Exposure Category, Internal Pressure Coefficient, and External Pressure Coefficient. The code utilizes all these properties and incorporates them with the shape of the building to locate the critical areas that will be affected by wind on a structure, and can be utilized to calculate the positive and negative pressures which will be experienced due to the wind. The structure is then

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designed to resist these values which have been calculated. Figure 2.1 has the zones of influence. 2.3 Different Types of Mini-Piles The types of mini-piles typically used in construction can be classified by three different systems. The first system of classification is by the method used for construction, and the second method is by the behavior of the piles. The third system of classification is the classification of piles by method of grouting. Classification by construction method gives a clear understanding of how each pile is made as well as the use of that specific design, Figure 2.2 has the different construction methods used. Pushed or driven piles are constructed by driving prefabricated piles into the ground either by hammering or through the use of hydraulic rams. These piles are often used to transfer light loads to the soil in a range from 3 to 30 tons. Compaction grouted piles are made by forcing the grout into the hole and generating a bulb on concrete at the base of the pile. These piles are excellent for the development of loads at shallow depths as the compaction increases the density of the soil which therefore increases its capacity. They are typically used for loads in the range of 15 to 75 tons. Jet grouted piles are created by filling the shaft with concrete traveling at a high velocity. This has the effect of greatly increasing the density of the soil far beyond the ability of other methods utilized while installing mini-piles. One benefit of jet grouting is that the capacity of the piles after construction can range from 50 to 150 tons.

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Post grouted piles are piles which have been modified after being cast in place. There is a void in the center of the shaft which runs connects from the tip to the surface, and concrete is then pumped under high pressure through this hole to increase the skin friction and the end bearing capacities of the piles in the range of 40 to 100 tons. Pressure grouted piles are constructed using concrete pumped into the shaft under high pressure. This pressure has the effect of increasing the density of the soil so that the pile is capable of generating larger resistive loads through skin friction from 25 to 75 tons. Finally, Drilled, End Bearing piles are piles constructed by drilling down to either bedrock or extremely dense soil, and then casting the pile in the hole which is generated. This type of shaft works by transferring loads directly to the tip of the pile and then into the soil, and does not rely on skin friction to resist significant loads. The capacity of Drilled, End Bearing piles ranges from 50 all the way to over 500 tons depending on the diameter of the pile and the material below the pile tip. Classification by behavior is based on the concept that piles will fall into only two categories, referred to as Cases, Figure 2.3 illustrates the various Cases. Case 1 refers to piles which directly resist loads which are applied on them, which is done either by an individual pile, or by a pile grouping. The loads will be applied axially and then be transferred to the soil. Case 2, by the classification based on behavior, is said to be of a Reticulated root pile structure. This form of pile behavior utilizes piles installed in a specific pattern for the purposes of confining the soil structure in the vicinity of the pile. The purposes of this can be for underpinning, stabilization, or earth retention.

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The final classification system is the classification of soil bases on the method utilized for the grouting process. Figure 2.4 is the classification based on the grouting process. Type A piles are those installed using concrete which is gravity fed into the hole. The pile is constructed either using a neat cement grout or a sand cement mortar. The piles are sometimes under-reamed at the base to aid the tensile performance. Type B piles are created by injecting neat cement grout into a hole as temporary steel drill casing, or the auger is removed. The pressures used for injection range from

43.5psi (0.3 Mpa) to 145psi (1 Mpa). The pressures uses are limited by the seal of the grout around the casing as it is being removed, and by the need to avoid hydrofracture pressures and excessive grout consumption. Type C piles are created by first installing a Type A pile with a grout pipe previously installed in the center. In the range of 15 to 25 minutes after the Type A pile is installed, neat cement grout of identical properties is then injected into the pile before the initial grout used has the ability to harden. The pressure used for this is generally about 145psi (1 Mpa). Type D piles are again constructed initially as Type A piles, and similarly to Type C, there is a grout pipe installed in the center. The difference between these two types, however, is that the Type D is pressurized several hours after the concrete has hardened, and the pressures utilized range from 290psi (2 Mpa) to 1,160psi (8 Mpa). A packer is also used in this method, and the reason for this is that if a specific area needs to be re-treated, this can be done several times without affecting the other horizons within the pile.

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For this project the Type A, Case 1' anchors were though to be the most economical while also providing adequate axial capacity to withstand wind-induced uplift forces via side shear resistance.

Figure 2.1 MWFRS Wind Influence Zones (Adapted from ASCE 7-02)

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Table 2.2 Design Pressures for Different Zones and SpeedsMain W ind Resisting System - Simplified Design Design W ind Pressures Exposure B at h = 30 ft. with I = 1.0 Zones Horizontal Pressures A 0-5 10 o 15 o 20 o 25 oo

Basic W ind Speed (mph)

Roof Loa Angles d (deg.) Case 1 1 1 1 1 2 30 - 45 o 0-5 10 o 15 o 20 o 25 oo

Vertical Pressures D -3.5 -3.1 -2.7 -2.3 2.4 ----7.0 7.0 -4.0 -3.5 -3.0 -2.6 2.7 ----7.9 7.9 -4.9 -4.3 -3.8 -3.2 3.3 ----9.8 9.8 -5.9 -5.2 -4.6 -3.9 4.0 ----11.8 11.8 -7.0 -6.2 -5.4 -4.6 4.7 ----14.0 14.0 E -13.8 -13.8 -13.8 -13.8 -6.4 -2.4 1.0 5.0 -15.4 -15.4 -15.4 -15.4 -7.2 -2.7 1.1 5.6 -19.1 -19.1 -19.1 -19.1 -8.8 -3.4 1.4 6.9 -23.1 -23.1 -23.1 -23.1 -10.7 -4.1 1.7 8.3 -27.4 -27.4 -27.4 -27.4 -12.7 -4.8 2.0 9.9 F -7.8 -8.4 -9.0 -9.6 -8.7 -4.7 -7.8 -3.9 -8.8 -9.4 -10.1 -10.7 -9.8 -5.3 -8.8 -4.3 -10.8 -11.6 -12.4 -13.3 -12.0 -6.6 -10.8 -5.3 -13.1 -14.1 -15.1 -16.0 -14.6 -7.9 -13.1 -6.5 -15.6 -16.8 -17.9 -19.1 -17.3 -9.4 -15.6 -7.7 G -9.6 -9.6 -9.6 -9.6 -4.6 -0.7 0.3 4.3 -10.7 -10.7 -10.7 -10.7 -5.2 -0.7 0.4 4.8 -13.3 -13.3 -13.3 -13.3 -6.4 -0.9 0.5 5.9 -16.0 -16.0 -16.0 -16.0 -7.7 -1.1 0.6 7.2 -19.1 -19.1 -19.1 -19.1 -9.2 -1.3 0.7 8.6 H -6.1 -6.5 -6.9 -7.3 -7.0 -3.0 -6.7 -2.8 -6.8 -7.2 -7.7 -8.1 -7.8 -3.4 -7.5 -3.1 -8.4 -8.9 -9.5 -10.1 -9.7 -4.2 -9.3 -3.8 -10.1 -10.8 -11.5 -12.2 -11.7 -5.1 -11.3 -4.6 -12.1 -12.9 -13.7 -14.5 -13.9 -6.0 -13.4 -5.5

Overhangs EOH -19.3 -19.3 -19.3 -19.3 -11.9 -----4.5 -4.5 -21.6 -21.6 -21.6 -21.6 -13.3 -----5.1 -5.1 -26.7 -26.7 -26.7 -26.7 -16.5 -----6.3 -6.3 -32.3 -32.3 -32.3 -32.3 -19.9 -----7.6 -7.6 -38.4 -38.4 -38.4 -38.4 -23.7 -----9.0 -9.0 GOH -15.1 -15.1 -15.1 -15.1 -10.1 -----5.2 -5.2 -16.9 -16.9 -16.9 -16.9 -11.4 -----5.8 -5.8 -20.9 -20.9 -20.9 -20.9 -14.0 -----7.2 -7.2 -25.3 -25.3 -25.3 -25.3 -17.0 -----8.7 -8.7 30.1 30.1 30.1 30.1 -20.2 -----10.3 -10.3

B -5.9 -5.4 -4.8 -4.2 2.3 ----8.8 8.8 -6.7 -6.0 -5.4 -4.7 2.6 ----9.9 9.9 -8.2 -7.4 -6.6 -5.8 3.2 ----12.2 12.2 -10.0 -9.0 -8.0 -7.0 3.9 ----14.8 14.8 -11.9 -10.7 -9.5 -8.3 4.6 ----17.6 17.6

C 7.6 8.6 9.6 10.6 10.4 ----10.2 10.2 8.5 9.6 10.7 11.9 11.7 ----11.5 11.5 10.5 11.9 13.3 14.6 14.4 ----14.2 14.2 12.7 14.4 16.0 17.7 17.4 ----17.2 17.2 15.1 17.1 19.1 21.1 20.7 ----20.4 20.4

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11.5 12.9 14.4 15.9 14.4 ----12.9 12.9 12.8 14.5 16.1 17.8 16.1 ----14.4 14.4 15.9 17.9 19.9 22.0 19.9 ----17.8 17.8 19.2 21.6 24.1 26.6 24.1 ----21.6 21.6 22.8 25.8 28.7 31.6 28.6 ----25.7 25.7

1 2 1 1 1 1 1 2 1 2 1 1 1 1 1 2 1 2 1 1 1 1 1 2 1 2 1 1 1 1 1 2 1 2

90

30 - 45 o 0 - 5o 10 o 15 o 20 o 25 o 30 - 45 o 0 - 5o 10 o 15 o 20 o 25 o 30 - 45 o 0 - 5o 10 o 15 o 20 o 25 o 30 - 45 o

100

110

120

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Table 2.2 (Continued)0 - 5o 10 o 15 o 20 o 25 o 30 - 45 o 0 - 5o 10 o 15 o 20 o 25 o 30 - 45 o 0 - 5o 10 o 15 o 20 o 25 o 30 - 45 o 0 - 5o 10 o 15 o 20 o 25 o 30 - 45 o 1 1 1 1 1 2 1 2 1 1 1 1 1 2 1 2 1 1 1 1 1 2 1 2 1 1 1 1 1 2 1 2 26.8 0.2 33.7 37.1 33.6 ----30.1 30.1 31.1 35.1 39.0 43.0 39.0 ----35.0 35.0 35.7 40.2 44.8 49.4 44.8 ----40.1 40.1 45.8 51.7 57.6 63.4 57.5 ----51.5 51.5 -13.9 -12.5 -11.2 -9.8 5.4 ----20.6 20.6 -16.1 -14.5 -12.9 11.4 6.3 ----23.9 23.9 -18.5 -16.7 -14.9 -13.0 7.2 ----27.4 27.4 -23.8 -21.4 -19.1 -16.7 9.3 ----35.2 35.2 17.8 20.1 22.4 24.7 24.3 ----24.0 24.0 20.6 23.3 26.0 28.7 28.2 ----27.8 27.8 23.7 26.8 29.8 32.9 32.4 ----31.9 31.9 30.4 34.4 38.3 42.3 41.6 ----41.0 41.0 -8.2 -7.3 -6.4 -5.4 5.5 ----16.5 16.5 -9.6 -8.5 -7.4 -6.3 6.4 ----19.1 19.1 -11.0 -9.7 -8.5 -7.2 7.4 ----22.0 22.0 -14.1 -12.5 -10.9 -9.3 9.5 ----28.2 28.2 -32.2 -32.2 -32.2 -32.2 -14.9 -5.7 2.3 11.6 -37.3 -37.3 -37.3 -37.3 -17.3 -6.6 2.7 13.4 -42.9 -42.9 -42.9 -42.9 -19.9 -7.5 3.1 15.4 -55.1 -55.1 -55.1 -55.1 -25.6 -9.7 4.0 19.8 -18.3 -19.7 -21.0 -22.4 -20.4 -11.1 -18.3 -9.0 -21.2 -22.8 -24.4 -26.0 -23.6 -12.8 -21.2 -10.5 -24.4 -26.2 -28.0 -29.8 -27.1 -14.7 -24.4 -12.0 -31.3 -33.6 -36.0 -38.3 -34.8 -18.9 -31.3 -15.4 -22.4 -22.4 -22.4 -22.4 -10.8 -1.5 0.8 10.0 -26.0 -26.0 -26.0 -26.0 -12.5 -1.8 0.9 11.7 -29.8 -29.8 -29.8 -29.8 -14.4 -2.1 1.0 13.4 -38.3 -38.3 -38.3 -38.3 -18.5 -2.6 1.3 17.2 -14.2 -15.1 -16.1 -17.0 -16.4 -7.1 -15.7 -6.4 -16.4 -17.5 -18.6 -19.7 -19.0 -8.2 18.2 -7.5 -18.9 -20.1 -21.4 -22.6 -21.8 -9.4 -20.9 -8.6 -24.2 -25.8 -27.5 -29.1 -28.0 -12.1 -26.9 -11.0 -45.1 -45.1 -45.1 -45.1 -27.8 -----10.6 -10.6 -52.3 -52.3 -52.3 -52.3 -32.3 -----12.3 -12.3 -60.0 -60.0 -60.0 -60.0 -37.0 -----14.1 -14.1 -77.1 -77.1 -77.1 -77.1 -47.6 -----18.1 -18.1 -35.3 -35.3 -35.3 -35.3 -23.7 -----12.1 -12.1 -40.9 -40.9 -40.9 -40.9 -27.5 -----14.0 -14.0 -47.0 -47.0 -47.0 -47.0 -31.6 -----16.1 -16.1 -60.4 -60.4 -60.4 -60.4 -40.5 -----20.7 -20.7

130

140

150

170

14

Figure 2.2 Classification Based on Method of Construction (Adapted from Hayward Baker Inc. 2003 PP18 [6])

Figure 2.3 Classification Based on Method of Grouting (Adapted from ISSMFE, TC-17 [7]) 15

Figure 2.4 Relative Relationship Between Mini-Pile Application, Design Concept and Construction Type (Adapted from ISSMFE TC-17)

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Chapter 3 Alternative Foundations 3.1 Determination of Forces on a 60x100x22ft-7in Building The structure used for the case study done in this thesis was a 60ft x 100ft steelframed building with a peak roof height of 22ft 7in. The structure was being relocated to 6308 44th Avenue East, Bradenton, Florida, and it was also being upgraded to the current building code which consisted of a higher design wind than when the building was originally constructed. While there were modifications done to the frame of the structure to resist these forces, the scope of this thesis will only consider how these forces will act on the foundation, and therefore all calculations done will be based on determining the wind load on the structure solely for the foundation design. The determination of forces was done in accordance with ASCE 7-02 (ASCE,2002) and the method used for calculations was the Main Wind Force-Resistance System (MWFRS). The building could have been designed using the simplified method according to the code, however the accuracy of these results would need some form of variation, and therefore both the simplified and the Analytical method were used to obtain the forces upon the structure. 3.1.1 Analysis Using the Simplified Procedure To use the simplified method, there were certain requirements which needed to be met by the structure. The first such requirement is that the structure be a simple diaphragm as determined in section 6.2 of the code. To be a simple diaphragm the code requires that the building be enclosed or partially enclosed with winds transmitted

17

through floor and/or roof diaphragms to the vertical MWFRS. Because this building consists of a steel frame, the forces acting upon the roof will then be transferred through the beams to the columns and from there directly into the foundation. For this reason, the structure can be classified as a simple diaphragm. The second requirement is that the building meets the classification in section 6.2 for a low rise building. The code stipulates that an enclosed or partially enclosed building having a mean roof height of less than 60ft, and that the mean roof height be less than the least horizontal dimension. This particular structure has a mean roof height of 21ft 3.5in and the least horizontal dimension is 60ft, therefore this requirement was also met. The third requirement is that the building is enclosed in accordance with section 6.2 and conforms to the wind-born debris provisions of section 6.5.9.3. An enclosed building, according to the code, is one which does not comply with the requirements for an open or partially enclosed building. An open building is one which has each wall being at least 80% open, a condition which this structure does not meet. A partially enclosed structure is one which has the total area of openings which receive positive external pressures being greater than the sum of the openings in the remainder of the structure by more than 10%, and that the total area of openings in a wall which receives positive external pressure exceeds 4ft2 or 1% of the area of the wall, whichever is smaller, and that the percentage of openings in the remainder of the building is less than 20%. The structure being designed does not contain any windows, and all of the doors are pull down shutters, therefore the conditions for a partially encloses structure are also not met, and therefore the building is termed as being enclosed.

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The code requires that a building be a regularly shaped building or structure, which is defined in the code as having no unusual geometrical irregularity or spatial form. Because this building is a simple rectangle, it can be classified as a regularly shaped structure. The simplified method also states that a building must not be classified as flexible in order for it to be used. Flexible structures are defines as slender structures with a natural frequency of less than 1Hz. Since this building is not slender, it can therefore not be classified as flexible. The sixth requirement for the simplified method is that the building does not have response characteristics making it subject to across-wind loading, vortex shedding, instability due to galloping or flutter; and does not have a site location for which channeling effects or buffeting in the wake of upwind obstructions warrant special consideration. This building does not have any specific response characteristics which would generate across-wind loading or vortex shedding. The building does not have any instability due to galloping or flutter, and is not located in an area where channeling effects or buffet requires any special consideration; therefore it also meets this requirement of the code. The seventh requirement is that the building structure has no expansion joints or separations. This structure contains no expansion joints or separations, therefore passes this requirement. The eighth requirement of the code is that the building is not subject to the topographical effects as described in section 6.5.7 of the code. This building is located

19

on level terrain with no nearby significant changes in elevation, and therefore is not subject to any topographical effects. The final requirement for utilization of the simplified method is that the building has a relatively symmetrical cross section in each direction and that the roof is flat or either hipped or gabled in nature with an angle of less than 45o. This particular structure has an angle of only 4.9 o, and therefore meets all of the qualifications for utilization of the simplified method. The first step in the simplified method is the determination of the basic wind speed in accordance with section 6.5.4. From this section the design wind speed for Bradenton, Florida was determined to be 130 mph. The importance factor for the structure was then determined from section 6.5.5 of the code. Since the building falls into Category I based on the classifications of table 1-1 of ASCE 7-02, and is in a hurricane prone region with wind velocities over 100mph, the importance factor I of the structure is 0.77. The exposure category for this structure is obtained from section 6.5.6 of the code. Because the structure is in a suburban area with closely spaced obstructions the size of single-family dwellings, the exposure category of this structure was determined to be category B. The height and exposure adjustment coefficient, 8 was then determined from Table 3.1. Because the mean roof height of the structure was 22ft 7in, and of exposure category B, Table 3.1 assigns all buildings under this category being less than 30ft in height an adjustment factor of 1.0; therefore the height and exposure adjustment factor for this building (8 ) is 1.0

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Table 3.1 Adjustment Factor for Building Height and Exposure, 8 Mean roof height (ft) Exposure B C D 15 1 1.21 1.47 20 1 1.29 1.55 25 1 1.35 1.61 30 1 1.4 1.66 35 1.05 1.45 1.7 40 1.09 1.49 1.74 45 1.12 1.53 1.78 50 1.16 1.56 1.81 55 1.19 1.59 1.84 60 1.22 1.62 1.87 The determination of the wind pressure for MWFRS is then done according to section 6.4.2.1 of the code, the formula is as follows: ps = 8 Ips30 where in this case 8 = 1.0, and I = 0.77. The value for ps30 horizontally across region A of the structure is 26.8psf, and likewise, the value longitudinally across region A is also 26.8psf. In the roof areas, the value for zone E was -32.2psf, and the value for zone B was -13.9psf (see Table 2.1 & Figure 2.1). After multiplying the values for ps30 by the exposure adjustment factor and the importance factor, it was determined that the walls of the structure would develop horizontal and vertical pressures both of 20.6psf, and that the roof would experience a pressure of -24.8psf in zone A, and -10.7psf in zone E. Table 3.2 shows the calculations for the simplified method. (Eq. 3.1)

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Table 3.2 Simplified Method SIMPLIFIED METHOD Wall Areas Horizontal A 26.8 psf Longitudinal A 26.8 psf Roof Areas Horizontal A -32.2 psf Horizontal B -13.9 psf Values Multiplied by importance Factor I= 0.77 Horizontal A 20.636 psf Longitudinal A 20.636 psf Roof Areas Horizontal A -24.794 psf Horizontal B -10.703 psf 3.1.2 Analysis Using the Analytical Method The next step in the process was to compare the simplified method to the Analytical Procedure as described in section ASCE 7-02 section 6.5. The first step in this procedure was to determine the basic wind speed V and directionality factor Kd in accordance with section 6.5.4 of the code. From these sections, V was determined to be 130mph, and Kd was determined to be 0.85. Step two was to determine the importance factor I of the structure in accordance with section 6.5.5. The importance factor for this structure was again determined to be 0.77. Step three was to determine the exposure category or categories and the velocity pressure exposure coefficients in accordance with section 6.5.6. The structure was determined to have category B exposure, and a velocity pressure exposure coefficient Kh of 0.7. The topographical factor Kzt was then determined from section 6.5.7 of the code. This value was determined to be 1. 22

Step five was to determine the gust effect factor in accordance with section 6.5.8 of the code. The determination of this gave a result for G as 0.85. Step six was to determine the enclosure classification of the structure, in accordance with section 6.5.9. The results of this, similar to those of the simplified method, are that the building is to be classified as an enclosed structure. Step seven was to determine the internal pressure coefficient GCpi in accordance with section 6.5.11.1 of the code. These were determined to be +0.18 and -0.18. The eighth step in the analysis was to obtain the external pressure coefficients GCpf in accordance with section 6.5.11.2. These were determined to be -0.43 in section 4e and 0.61 in section 1e for the walls; with -1.07 in section 2e and -0.53 in section 3e of the roof. The ninth step was the determination of the velocity pressure in accordance with section 6.5.10 in the code. The formula for this is as follows: qz = 0.00256 KzKztKdV2I (psf) (Eq. 3.2)

The worst case pressures exerted on the wall based on the calculations were +15.7psf and -12.1psf in the transverse direction. Longitudinally, the worse case loads were determined to be +15.7psf and -12.1psf. The roof, however experienced negative pressures in all cases, and in both the transverse and longitudinal direction the value of the negative pressure was determined to be -24.8psf, which is the same value determine by utilizing the simplified method. Table 3.3 shows the values obtained using the analytical method.

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Table 3.3 Analytical MethodDESIGN PROCEDURE 1) From Section 6.5.4 Determine basic wind Speed V Determine Directionality Factor Kd 2)From Section 6.5.5 Determine the Building Category Determine the Importance Factor I 3)From Section 6.5.6 Determine Exposure Category Determine Kz or Kh 4) Determine topographic factor Kzt 5) Determine the Gust Effect Factor G 6) Determine the enclosure classification 7) Determine the Internal Pressure coefficient Gcpi 130 0.85 I 0.77 B 0.7 1 0.85 ENCLOSED 0.18 -0.18 mph

8) Determine the External Pressure Coefficient Gpf 4e -0.43 1e 0.61 Roof 2e -1.07 3e -0.53 determine qz, qz = 0.00256 Kz Kzt Kd V^2 I = 0.00256 * 0.7 * 1 * 0.85 * (130 * 130) * 0.77 19.821402 qh = qz p = qh (GCpf) - Gcpi) Transverse Wall worse case p = qh (1e - (+/-) 0.18) positive on wall 15.658907 psf for wall worst case p = qh (4e - (+/-) 0.18) negative on wall -12.091055 psf for wall Longitudinal Wall worse case p = qh (1e - (+/-) 0.18) positive on wall 15.658907 psf for wall worst case p = qh (4e - (+/-) 0.18) negative on wall -12.091055 psf for wall Transverse Roof worst case p = qh (2e - (+/-) 0.18) roof uplift 1 -24.776752 psf worst case P = qh(3e - (+/-) 0.18) roof uplift 2 -14.073195 psf USE WORST CASE, ASSUME 24.77 PSF ON ROOF Longitudinal Roof worst case p = qh (2e - (+/-) 0.18) roof uplift 1 -24.776752 psf worst case P = qh(3e - (+/-) 0.18) roof uplift 2 -14.073195 psf

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3.1.3 Determination of Dead Load The determination of the dead loads of the structure began with the calculation of the roof load over the column with the largest tributary area. This area was right above the interior columns of the structure. The area of load which each roof column would receive was determined to be 25ft x 30 ft, or 750ft2. The weight of the roof over this area contributed a load of 0.75kips to the foundation of the structure. The second step in the analysis was to determine the weight of the purlins which support the roof. The purlins used were L7.5x3.75x0.125 with a weight of 4.57plf. These purlins were also spread over the tributary area, and it was determined that there would be 8 purlins each 25ft long which would contribute to the load on the foundation. The weight of these turned out to be approximately 0.92kips. The girders used were W8x10, and the tributary length of each girder was 30ft long. These girders have a weight of 10plf, and therefore the weight of the girders was 0.3kips. The weight of the columns was then determined. The columns used were W8x21, and were 20ft in length. With a weight of 21plf, the weight of the columns was determined to be 0.42 kips. This structure also contained side purlins connected to the columns. These were 8x2.5 Z, with a weight of 4.95plf. These purlins had a tributary width of 25ft, and each column had 4 purlins mounted on it. The weight of these purlins was determined to be approximately 0.50kips. The total weight of the frame of the building being analyzed

25

was determined to be 2.89 kips, based on the calculations. Table 3.4 is the calculation of the dead load. Table 3.4 Dead Load Calculation SELF WEIGHT ROOF WEIGHT Tributary Width Tributary Length Per Wall Tributary Roof Area Weight of 26 gauge steel siding Roof Weight

25 30 750 1

ft ft ft^2 psf

0.75 kips

FRAME WEIGHT Top Purlins (L 7.5 x 3.75 x 0.125) Number of Purlins 8 Tributary Width of Purlins 25 ft Weight of Purlin 4.57 plf Purlin Weight 0.914 kips Girder ( W8x10) Tributary Length 30 ft Weight 10 plf Girder Weight 0.3 kips Column (W8x21) Length 20 ft Weight 21 plf Column Weight 0.42 kips Side Purlins Number of Purlins 4 Tributary Width of Purlin 25 ft Weight 4.95 plf Side Purlin Weight TOTAL WEIGHT 0.495 kips 2.879 kips

3.1.4 Determination of Uplift Force to be Resisted From the calculations done in section 3.1.1 and 3.1.2, it was concluded that the uplift force acting on the building would be 24.8psf. When this load is multiplied over 750ft2, which is the tributary roof area over the interior columns, an uplift force of 18.6kips obtained. This 18.6kips is the upward force which will be exerted on the

26

structure during a hurricane with wind speeds of 130 miles per hour. At this point, the only resistance which exists to this large uplift force is the self weight of the structure. This self weight was calculated as 2.89kips. The net uplift force which needed to be restrained was therefore the difference between the total uplift force and the self weight of the building. This net uplift force was therefore 15.71kips, however, the designers specified that the foundation be required to resist an uplift force of 17kips. 3.2 Incorporation of Both Tension and Compression Forces in Footing Design The tensile forces in this structure are so large that they govern that design. For this reason, the initial analysis of mini-piles vs bulk footings shall be focused on designing the foundation to resist these forces. 3.2.1 Design of a Mass Concrete Footing to Resist Uplift Forces The first step in designing the footing to resist the forces is the determination of the factored load to be resisted. The safety factor incorporated in the foundation design of this structure using a mass concrete footing is 1.5, and the code allows only 80% of the dead load to be utilized to resist uplift forces, therefore the load to be resisted is 32kips, which is 32,000 pounds. The unit weight of concrete is 150pcf, and therefore the total volume of concrete needed to resist the uplift forces in this structure will be 213.4 cubic feet per column. The 213.4 cubic feet of concrete required to resist the load can be constructed using a bulk footing at the base of each column. If this is done, the footing required would be six feet deep, and have a cross sectional area of 36ft2 (6ft x 6ft x 6ft). Considering that this volume of concrete is required for just one footing, the total volume

27

of concrete needed to resist the 140 kips of uplift for the entire structure was calculated to be 1750 cubic feet. 3.2.2 Design of Mini-Piles The first step in the design of the mini-piles was a site evaluation. Cone Penetration Tests (CPT) were performed at each of the four corners of the proposed foundation in accordance with ASTM D-3441 (ASTM, 1996) [8] see Figure 3.1. Figure 3.2 shows the results from the NE sounding. Using correlations developed by Robertson and Campanella (1983)[10], the tip stress and friction ratio were used to identify the soil type from 12 pre-defined regions in Figure 3.3. These classifications also help to convert the CPT data to equivalent Standard Penetration Test resistance values, (N) also shown in Figure 3.2. Figure 3.4 shows the CPT data plotted on the Robertson & Campanellas classification chart and shows mostly low friction ratio (cohesion less) soils. Figure 3.4 shows the values converted to soil type. Similar results for all four CPT soundings can be found in the Appendix (Fig B-5 through B-8) along with interpreted results. With the soil stratification and strength identified from the CPT data, a spreadsheet was designed to determine the capacity of the shaft as a function of its length using both the CPT method and the Titanmethod. 3.2.2.1 CPT Method Design using the CPT method utilizes the side shear forces directly measured from the CPT tests to determine the capacity of the min-pile. The diameter of the anchor is determined prior to construction, and in this case was 6in. From this, the perimeter

28

was calculated to be 1.57ft. The perimeter times the length is the area in contact with the soil, and is therefore responsible for the side shear which develops. Knowing that the shaft has a perimeter of 1.57ft2/ linear ft, we can multiply this value by the side shear determined by the CPT test to develop the skin friction of the mini-pile at a given depth. By adding up the capacity of the shaft up to a given depth, the capacity of that depth can be determined. This was done to develop the capacity of the shaft up to the maximum depth of which the CPT machine was able to achieve (between 17ft and 19ft). The total capacity of the shaft for a given depth was calculated on the same spreadsheet, and then a VLOOKUP function was used to obtain the minimum depth required to resist the factored uplift force. The CPT method determined this minimum depth to be 16.4ft. for the shaft with the weakest soil strata. Table 3.5 is the minimum required length for each pile using the CPT method. Table 3.5 Minimum Pile Length Using The CPT Method USING CPT DATA Design Lengths Uplift Compression MINIMUM DESIGN LENGTH (ft) NE Corner 16.47 9.43 16.47 NW Corner 10.86 6.03 10.86 SE Corner 15.78 8.76 15.78 SW Corner 12.48 6.57 12.48 3.2.2.2 Titan Method The Titan method for the calculation of minimum shaft length takes a different approach to determining the depth necessary to resist the uplift forces required. The first step is to determine the nominal diameter of the shaft being drilled. As before, the diameter is 6in Next the ultimate factored load was determined which the system must resist. This load (Qu) was 17kips. A value for the shear resistance of the soil (qsk) was then

29

determined based on its classification. This particular site had 15ft of sand to silty sand, followed by at least 2ft of very stiff clay or clayey silts, with limestone being below that. Because of this, a value for the density was determined to be 91.8lbs/ft3. The soil type is then also used to determine the grout body factor of the shaft. The grout body factor is said to be the true diameter of the shaft based on the expansion of the concrete into the surrounding soils, therefore there is an amplification in the diameter and surface area of a mini-pile specific to each soil type. For this particular soil, the grout body factor is 1.5, which produced a grout body diameter (d) of 0.22m. The Titan method then utilizes a formula to calculate the length of the shaft. The formula which is used is the following: L = Qu / (B x d x qsk) (Eq. 3.3)

The calculations done based on this formula generated a minimum required length of 9.2 ft. Table 3.6 is the minimum pile length using the Titan method. Table 3.6 Minimum Pile Length for NE Corner Using The Titan Method USING TITAN METHOD d= 0.1524 m Qu 151.239534 KN qsk 150 KN/m^3 Grout Body Factor 1.5 Grout Body Diameter 0.2286 m Required Pile Length 2.80675 m Required Pile Length 9.2 ft 3.2.3 Compression in the Footing The maximum compressive force which the foundation is expected to resist was determined to be 10kips. When combined with the end bearing of the mini-pile, the capacity in compression is far greater than 17 kips, therefore no special modification is needed for the structure to resist the 10kip compression load. As for the pressure placed

30

on the concrete, the area of the shaft is 0.78ft, and when the 10 kips is distributed over this area, there is a compressive force of 12, 800psf, which is only 88.4psi, which can easily be resisted by the concrete. 3.3 How Testing Can Aid in Safety and Economy 3.3.1 Safety Testing can play an important role in quality assurance, especially in the design of foundations. This is because unless adequate testing is done, there is no way of knowing exactly what the properties of the soil are beneath the surface. If the design is done using strictly a bulk footing, then there is no need for testing, as the only force responsible for the resistance of the forces is gravity. For mini-piles, however testing should be used to confirm soil shear strength values. The Titan method recommended a shaft length of only 9.2ft to resist 17kips of uplift force. This was based on a lot of assumptions. One assumption is that the effective diameter of the shaft was in fact 1.5 times that of the nominal bore hole diameter. Without testing, a shaft could be constructed based on the assumption that the soil is relatively consistent in nature, and this could lead to a structural failure when full design loads are realized.. The design using the CPT method begins with an evaluation of the site using the CPT data. This gives you the exact profile of the soil beneath the surface, and the exact tested value for the skin friction which the soil is capable of generating. By using the CPT method, each shaft is built to meet the characteristics of a specific location. This leads to a much safer design, and testing which is discussed in chapter 4 confirms that the CPT method is conservative in its values for ultimate capacity.

31

Both methods are empirical and cannot be expected to fully predict the exact side shear development in the anchors/mini-piles. As a result, testing provides a means to confirm design assumptions prior to full construction. 3.3.2 Economy As safety, economy, and uncertainty are all linked in the design of foundations, testing provides a means by which to eliminate uncertainty and help assign reasonable safety factors. Higher safety factors cause higher costs and vice versa. Therein, a no testing approach typically employs safety factors no less than 3.0; whereas testing programs have associated safety factors no greater than 2.0. As the frequency of testing increases to 100% verification, the safety factor can fall as low as 1.0. Mini-piles are easily tested and it is not unreasonable to test every anchor/mini-pile. As the safety factor is directly related to anchor lengths and the associated cost, testing mini-pile anchors can lead to cost savings ranging from 50% to 200% and above, while reducing uncertainty to near zero. P Service Load f > 0% Where the Safety Factor ranges from 1 to 3 for testing frequency of 100 to 0% respectively. (Eq. 3.4)

32

Figure 3.1 USF CPT Truck

Figure 3.2 Results from the NE CPT Sounding

33

Figure 3.3 Soil Classification Chart (Adapted from Robertson & Campanella)

Figure 3.4 Soil Classification for NE Corner

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Chapter 4 Construction and Testing 4.1 Site Investigation Prior to the design and construction of the mini-piles for this project, a site investigation was done to determine the soil characteristics. On October 18th, 2004 Cone Penetration Tests were performed at the locations of the 4 corners of the proposed foundation, in accordance with ASTM D-3441. The results of the test were that the soil below the foundation consisted of 15ft of sand to silty sand with 2 to 5ft of very stiff clay to clayey silts beneath that. Between 17ft and 19ft penetration refusal was encountered by the CPT machine, indicating that this was the top of the limestone. The data from these tests was then used to determine the worst case minimum shaft length of the minipiles. This concluded that the worst case minimum shaft length needed was 17ft, and the designers recommended that the shafts therefore be extended to a depth of 20ft for added safety, based on the fact that the greatest cost involved in the installation of the mini piles is the mobilization, and therefore adding three feet onto each shaft would generate a significantly stronger pile for only a small increase in cost. 4.2 Field Test On October 28, 2005 an out of position test mini-pile was installed 10 ft. east of the proposed SE corner. This test pile was drilled using a 4in diameter bit, and had a nominal design diameter of six inches (when in sand). The embedment length was 22ft 3in, and the overall anchor length was 26ft 3in as shown in Figure 4.1. A single Williams Form Bar was placed in the center of the shaft, with a tensile strength of 150ksi, shown in

35

Figure 4.2. The borehole was made using a CME-45 drill rig typically used for performing SPT tests. While the hole was being drilled, a weak concrete mixture consisting of a 0.88 water to cement ratio was used as the drilling fluid in order to prevent the sides from collapsing and to flush soil debris to the surface. At the desired drilling depth, a stronger mixture consisting of a 0.45 water to cement ratio was pumped through the drill stem into the hole, which forced the weaker/less dense concrete to rise to the surface. Following this, the drill stem was removed, and the threaded anchor bar was placed into the hole. A portion of the bar was left exposed at the top, for the purpose of testing. Figure 4.3 shows the anchor installation process. On November 2, 2005 the capacity verification test was performed in accordance with ASTM D-1143 (ASTM 1996), Figure 4.4 shows the test setup. The first of the two load cycles which were performed generated an uplift force of 43kips, while the second one went up to 64kips, Figure 4.5 shows the load test results. These tests resulted in no significant permanent deformation of the shaft, and there were no indications that the pile was close to failure, even though it was stressed to almost four times the design load. The testing also concluded that at 17kips, the upward displacement of the pile was only 1/8in, concluding that the mini-piles were more than adequate to resist the design loads. 4.3 Site Survey Following the confirmation that the mini-pile design was adequate, the site survey was done to mark off the footprint of the building. The survey started by marking off the 4 corners of the building, and once this was established, the locations for all 16 mini-piles were marked off. Figure 4.6 shows the site plan and location of all columns/anchors.

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The next step was to ensure that all of the piles would end at the exact same elevation, to ensure that they would all have the same relative height to the top of the slab, thereby ensuring proper load distribution. The first step in this process was to determine the highest point of the ground within the footprint of the area, and have that be the benchmark for the site. From there, the height of the slab was determined relative to the ground. The tops of the piles were required to be 20in below the top of the slab, so notes were made for the adjustment of each pile so that they could all end at the same height, relative to the finished slab elevation. 4.4 Mini-Pile Installation After the survey was completed, the next step was to install the piles. The installation was doing using a water truck, to provide the necessary water for the grout, and a SPT drill rig, to drill the holes and pump the concrete. Figure 4.7 is the SPT drill rig. The installation began by first positioning the drill rig over the position of each hole (accurate to within one inch). The second step was to install the mud pan, which is used to re-circulate the grout, around the hole to ensure there is a watertight seal between the pan and the earths surface. After the pan was secure, an initial amount of cement with a water to cement ratio of 0.88 was poured into the pan and allowed to circulate through the pump mounted onto the SPT machine. After grout circulation was established, the drilling commenced. The drill rods used for construction were each 5ft long, and therefore 4 bars had to be used to drill down 20ft from the surface of the earth. The level of grout in the mud pan was monitored, as less grout would return to the surface because of the increased volume of the hole, and when the volume of the pan was

37

low, more cement and water of the same consistency was added to compensate for the change in volume. Once the drill bit reached the desired depth of 20ft, a stronger concrete grout consisting of a 0.45 water to cement ratio was pumped into the mud pan and then circulated into the hole. The weaker cement grout was discarded as it rose to the surface, and pumping continued until the stronger grout had filled the entire volume of the mini-pile as evidenced by the change in slurry color (from greyish to greenish). At this point, the removal of the drilling rods began. While the rods were being lifted out of the ground, grout was continuously pumped into the hole through the tip of the drill bit to ensure that there would be no voids left by the removal process. When the final rod, containing the drill head, was removed; the steel bar was installed into the shaft. The bar used for these mini-piles was a #7 reinforcement bar, with a yield capacity of 60ksi. The area of a #7 bar is 0.6in2, therefore the maximum tensile force will be only 28ksi for the designed uplift force, making the bar acceptable even with the specified safety factor of 2 for the substructure design. The bar was picked up by the boom on the drill rig, and hoisted until it was perfectly vertical. It was then centered over the hole, and slowly lowered to ensure that it did not enter the shaft at an angle, and centralizing tabs were attached to the bar at the top and bottom to assure adequate cover, shown in Figure 4.8. Once in, the hook on the top of the reinforcing bar was set to the right height, as determined during the site survey, and then the pile was left to harden. After the reinforcement was placed in the ground, the remaining grout was discarded, and the seal of the mud pan with the ground was broken. The equipment was relocated to the site of the next mini-pile, and the process was repeated until all piles

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were completed. On average, it required about 1 hour to relocate the equipment, drill a hole, and install the anchor rod. Following the hardening of the concrete, the formwork was then put in place for the construction of the slab. The area around each mini-pile was dug out so that a special connection could be made to ensure that there would be a proper transfer of force from the columns to the piles, this connection is shown in Figure 4.9. After this was done, all of the reinforcement was placed and tied, and then the entire slab was poured in a single monolithic fashion.

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Figure 4.1 Installed Mini-Pile

Figure 4.2 Williams Form Bar 40

Figure 4.3 Anchor Installation

41

Figure 4.4 Static Load Test Setup

Figure 4.5 Load Test Results

42

Figure 4.6 Site Layout (Courtesy of Structural Engineering and Inspections, Tampa)

43

Figure 4.7 SPT Drill Rig

44

Figure 4.8 Centralization Tabs

Figure 4.9 Load Transfer Connection

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Chapter 5 Economy of Foundations 5.1 Mass Concrete Footer Following the calculation of the uplift force which the footing must resist, a volume of concrete required for the mass concrete footer was determined. The cost of these footers is both a function of the volume of concrete poured, and the dimensions of the footer. It was assumed that the footer resisting the 17 kips of force would require a footing with a volume of 216 cubic ft This generated values which could be used to estimate the total cost of the project. Table 5.1 is the estimated construction cost by item. Table 5.1 Construction Cost by Item ESTIMATION OF FOOTER CONSTRUCTION COSTS Pour Concrete 174 $/CY 6 CY/MH Rebar Weight/CY of Concrete 172 LB/CY Purchase Cost 0.8 $/LB Tie In Place 0.13 $/LB Labor $15 $/HR The dimensions of the footer were used to also generate estimated values for the labor costs required to produce each footer. These values, combined with the cost of the raw materials were then used to produce a cost for each footer to be built. The estimated cost of each footer in the project was determined to be $2,691.68. To resist the entire 140 kips required in this design would therefore cost $21, 807.59. Table 5.2 is the cost per footer, and Table 5.3 is the total cost of the project using a mass concrete footing.

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Table 5.2 Cost for Each 17 Kip Mass Concrete Footer MASS CONCRETE FOOTER Footer Volume 216.00 ft^3 8.00 yd^3 Footer Length 6.00 ft Footer Width 6.00 ft Footer Height 6.00 ft SET REINFORCEMENT Reinforcement weight 1376.00 Lbs Reinforcement Cost 1100.80 $ Labor Cost 178.88 $ POUR CONCRETE Concrete Cost 1392.00 $ Labor 1.33 MH Labor Cost 20.00 $ TOTAL COST $ 2,691.68 Table 5.3 Total Cost Using Mass Concrete Footer MASS CONCRETE FOOTER Footer Volume 1750.00 ft^3 64.81 yd^3 Footer Length 6.00 ft Footer Width 6.00 ft Footer Height 48.61 ft SET REINFORCEMENT Reinforcement weight 11148.15 lbs Reinforcement Cost 8918.52 $ Labor Cost 1449.26 $ POUR CONCRETE Concrete Cost 11277.78 $ Labor 10.80 MH Labor Cost 162.04 $ TOTAL COST $ 21,807.59 5.2 Mini-Pile Anchor The determination of the cost of each footer using mini-piles differs from that of the mass concrete footing. This is because there is no formwork or excavation, and the volume of material used is measured in bags of cement rather than cubic yards of concrete. Also, there is no labor cost per activity, and instead there is an initial mobilization cost and an operational cost per foot drilled. To determine the cost per

47

min-pile in this project, the mobilization cost was divided by the number of piles to generate the cost per pile. Based on the same assumptions made for the calculation of the mass concrete footer, the material costs were calculated to determine the cost per pile. With each pile receiving an equal share of the mobilization cost, it was determined that the cost per pile resisting 17 kips was $1,723.50. This value is 64% of that for the mass concrete footer. The project cost after designing each footing based on the NE (worst case) sounding and drilling to the minimum required length is therefore $15,624.00, which is a savings of 28% of the original foundation cost. Table 5.4 is the cost per 17 kip mini-pile, and Table 5.5 is the total cost of the project using mini-piles. Table 5.4 Cost for Each 17kip Mini-Pile MINI-PILE INSTALLATION Mobilization 5000 $/Project Operation 80 $/ft Concrete 0.5 Bags/ft 6 $/Bag # Piles 16 Pile Length 17 ft Mobilization cost 312.5 $/Pile Labor 1360 $/Pile Concrete Used 8.5 Bags Concrete Cost 51 $/Pile TOTAL COST $1,723.50 $/Pile Table 5.5 Total Cost Using Mini-Piles Mini-Pile Installation Cost Total Drill Length 128 Mobilization 5,000.00 Operation 10,240.00 Concrete 64 384 TOTAL COST $15,624.00

ft $ $ Bags $

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5.3 Break Even Analysis From the costs calculated for both types of footings, it is obvious that it is more economical to use mini-piles for large projects, however the mobilization costs associated with constructing the mini-piles makes it possible that there will not always be instances where the mini-piles are a cheaper alternative to mass concrete footings. This is because the cost of one 17 kip mass concrete footing is $2,691.68, while the cost of mobilization alone for the equipment to install the mini-piles is $5,000. A break even analysis is therefore necessary to determine if there is a specific force which must be resisted at which one option is more economical than the other. The first step to determine this was to determine the cost per kip resisted by the mass concrete footer. This was done by dividing the total cost of the project by the force resisted. This gave the mass concrete footer a value of $155.77 per kip resisted. This value is linear from the origin, as a force of 0 kips will require $0. Table 5.6 shows the footer cost per kip resisted as well as that for the mini-piles. Table 5.6 Cost Per Kip Resisted COST PER KIP RESISTED MASS CONCRETE Total Load Resisted 140 Cost of Mass Concrete Footing 21,807.59 Cost Per Kip Resisted 155.77 MINI-PILE Total Length of Mini-piles 128 Cost Per Ft. Drilled 80.00 Total Drilling Cost 10,240.00 Total Concrete Cost 384 Cost Per Kip Resisted (W/O Mob.) 75.89

kips $ $/Kip ft $ $ $ $/Kip

The determination of the cost per kip resisted by the mini-piles differs from that of the concrete footing due to the mobilization cost. To do the analysis of cost per kip

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resisted, the mobilization cost was subtracted from the total cost of the project. The remaining value was then divided by the total force resisted, and this generated a value of $75.89 per kip resisted. The reason why the mobilization cost was subtracted before the determination was done was that this cost is constant and must be paid regardless of the force resisted (or number of anchors installed). The calculation of the cost per kip resisted by the mini-pile system is therefore the addition of the mobilization cost with the product of the cost per kip resisted and the number of kips resisted. When plotted, this value is also linear, however does not begin from the origin of the graph, but rather is shifted up by $5000. The results of the plots show that the break even point for this specific project, based on the soil conditions, the break even force for the design is going to be 63.59kips, at a cost of $9,749.68. Figure 5.1 is the results of the break even analysis. After the break even analysis was done, an attempt was made to determine if there are any common soil types for which the used of mini-piles would not be economical. This was done by determining the capacity of mini-piles in soils with typical properties. As no in-situ values are available for this, the CPT method could not be utilized for this analysis, and therefore the Titan method was used to generate the graph for Figure 5.2. The data generated was then used to plot a graph of the cost of a foundation as a function of the required resistive load, shown in Figure 5.3. This graph has a different break even point from the case study, because the CPT Method is more conservative than the Titan method, however it can clearly be seen that a harder/denser soils can resist loads for a lower price than would be required for the same type of foundation constructed in a softer soil. This graph is based on certain soil assumptions,

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and can therefore only be used a proof that there is a point for all soils at which a mass concrete footing would be a less economical option to that of a mini-pile. One significant characteristic of the graph is that regardless of the soil type, the mass concrete footer is the more economical choice for resisting light loads, and the savings in foundation design can only be realized when designing for the resistance of large uplift forces.

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Figure 5.1 Soil/Site Specific Break Even Analysis

Figure 5.2 Mini-Pile Anchor Capacity Based on Soil Type (Titan Method)

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Figure 5.3 Foundation Cost Per Force Resisted (Using Titan Method)

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Chapter 6 Conclusion and Summary The Conclusion of this thesis is that the use of mini-piles can be economical in any soil type provided the force required to resist is larger than the break even point specific to that soil. This break even point is site specific as it is will be determined solely by the soil in that specific location. Because of this, a mini-pile foundation might be more economical for resisting a given load in one spot, and not in another because of the difference in the soil at the different sites. However, it was also concluded that there is a break even point associated with each soil type, and therefore there will be a point at which a foundation will be more economical were it constructed using mini-piles vs using a mass concrete footer. Not mentioned in this study were savings associated with the time required for foundation construction. Had a mass concrete footer been utilized, there would have been a certain time required to excavate the foundation and fabricate the reinforcement cage needed for the footers. Mini-piles, however, do not require any excavation prior to installation, and therefore saves time in construction. For the case study, a crew of 4 men installed all of the anchors in 2 days, while casting of the mass concrete footer would have taken 4. To summarize, mini-piles are very economical for resisting uplift forces, as once the mobilization cost is recovered, the cost per kip resisted is far less than that of a mass concrete footer. There are also foreseeable benefits to increased utilization of mini-pile anchors. This is because the cost of a foundation constructed using mini-piles is affected

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less by the increase in the cost of cement, than that of a mass concrete footer, which will make mini-pile anchors a much more economical option as the price of cement increases.

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References [1] ASCE 7-02 Minimum Design Loads for Buildings and Other Structures, Reston, Virginia, 2003. Gunaratne, M. The Foundation Engineering Handbook. CRC Press. 2006. Drilled and pressure grouted TITAN Micro Piles, The Con-Tech Systems Ltd. Website. Last Accessed on 27 October 2006. USGS Mineral Cost of Cement, The United States Geological Survey Minerals Information Website. Last Accessed on 27 October 2006. Manual of Steel Construction Load and Resistance Factor Design. Prepared by the American Institute of Steel Construction Inc. 2003. MICROPILES Project Support From the Ground Down, The Hayward Baker Geotechnical Construction Website. Last Accessed on 27 October 2006. Micropiles, The International Society for Soil Mechanics and Geotechnical Engineering Website. Last Accessed on 27 October 2006. Standard Test Method for Deep, Quasi-Static, Cone and Friction-Cone Penetration tests of Soil, D3441-94. Annual Book of ASTM Standards, Volume 04.08, American Society for Testing and Materials. 1996. Standard Test Method for Piles Under Static Axial Compressive Load, D1143-81. Annual Book of ASTM Standards, Volume 04.08, American Society for Testing and materials. 1996. Robertson, P.K. and Campanella, R.G., Interpretation of Cone Penetration Tests. Part I: Sand, Canadian Geotechnical Journal, Vol. 20, No. 4, Nov. 1983, pp. 718 - 733.

[2] [3]

[4]

[5]

[6]

[7]

[8]

[9]

[10]

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Appendices

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Appendix A Mini-Pile Location and Structural Plans of Case Study In this appendix is the site layout with the location of each mini-pile which was constructed. Also included are the structural plans for the building.

Figure A.1 Site Layout

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Appendix A (Continued)

Figure A.2 Roof Framing Plan

59

Appendix A (Continued)

Figure A.3 Front and Rear Elevations

60

Appendix A (Continued)

Figure A.4 Side Elevations

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Appendix B Results of CPT Testing This appendix shows the graphical and digital results for the CPT tests which were performed for this project.

Figure B.1 CPT Sounding for NE Corner

Figure B.2 CPT Sounding for NW Corner

62

Appendix B (Continued)

Figure B.3 CPT Sounding for SE Corner

Figure B.4 CPT Sounding for SW Corner

63

Appendix B (Continued)

Figure B.5 Soil Classification for NE Corner

Figure B.6 Soil Classification for NW Corner

64

Appendix B (Continued)

Figure B.7 Soil Classification for SE Corner

Figure B.8 Soil Classification for SW Corner 65

Appendix B (Continued) Table B.1 CPT Sounding for NE CornerDepth (ft) 0 0.01 0.16 0.36 0.55 0.75 0.95 1.13 1.3 1.49 1.67 1.84 2.02 2.21 2.39 2.58 2.77 2.97 3.17 3.37 3.56 3.77 3.97 4.17 4.37 4.57 4.78 4.98 5.17 5.38 5.58 5.78 5.98 6.19 6.39 6.59 6.78 6.98 7.18 7.36 7.55 7.74 7.93 8.12 8.3 8.49 8.68 8.86 9.05 9.25 9.43 9.62 9.8 9.99 10.19 10.37 Soil Type No Reading Sandy Silt to Clayey Silt Silty Sand to Sandy Silt Silty Sand to Sandy Silt Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand Sand Sand Sand Sand Sand Sand Sand to Silty Sand Silty Sand to Sandy Silt Silty Sand to Sandy Silt Silty Sand to Sandy Silt Silty Sand to Sandy Silt Sandy Silt to Clayey Silt Sensitive Fine Grained Sensitive Fine Grained Sensitive Fine Grained Sensitive Fine Grained Sensitive Fine Grained Sensitive Fine Grained Sensitive Fine Grained Sensitive Fine Grained Sandy Silt to Clayey Silt Silty Sand to Sandy Silt Sandy Silt to Clayey Silt Sensitive Fine Grained Sensitive Fine Grained Sensitive Fine Grained Sandy Silt to Clayey Silt Silty Sand to Sandy Silt Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand N Value n/a 8 12 11 10 13 19 19 21 22 23 24 23 19 20 13 11 9 7 7 10 6 4 4 4 4 5 8 9 8 8 9 7 10 6 8 11 13 13 15 13 11 11 13 15 13 13 11 12 12 12 13 13 13 12 14 Tip Stress (bars) 8.94 17.48 37.35 32.52 41.57 53.59 77.4 96.32 104.98 114.14 115.53 120.94 119.08 95.38 83.09 40.24 33.44 28.94 22.62 14.99 10.43 5.63 4 3.77 3.94 3.89 4.52 7.77 18.7 25.61 16.35 8.52 7.22 10.07 12.57 25.7 43.85 52.84 52.64 61.19 53.03 45.51 47.45 52.12 59.94 54.59 53.28 47.4 48.62 47.73 51.17 52.09 52.48 55.17 49.45 59.3 Sleeve Stress (bars) 0 0.07 0.68 0.18 0.14 0.16 0.21 0.31 0.35 0.48 0.59 0.53 0.6 0.56 0.6 0.61 0.51 0.22 0.08 0.06 0.05 0.05 0.04 0.03 0.03 0.02 0.04 0.03 0.05 0.04 0.06 0.01 0.02 0.01 0.03 0.07 0.14 0.18 0.18 0.24 0.23 0.17 0.19 0.16 0.19 0.2 0.16 0.21 0.18 0.22 0.12 0.16 0.23 0.23 0.18 0.2 Friction Ratio (%) -0.06 0.39 1.83 0.56 0.35 0.29 0.28 0.32 0.34 0.42 0.51 0.44 0.5 0.59 0.73 1.51 1.51 0.77 0.33 0.38 0.5 0.88 0.92 0.81 0.72 0.64 0.78 0.41 0.26 0.17 0.38 0.08 0.25 0.13 0.22 0.26 0.32 0.35 0.33 0.39 0.44 0.37 0.4 0.3 0.31 0.36 0.3 0.45 0.37 0.46 0.23 0.31 0.44 0.42 0.35 0.34

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Appendix B (Continued) Table B.1 (Continued)10.54 10.73 10.92 11.1 11.29 11.48 11.67 11.85 12.03 12.22 12.41 12.59 12.77 12.96 13.15 13.33 13.51 13.71 13.89 14.07 14.25 14.43 14.62 14.8 14.97 15.16 15.35 15.53 15.72 15.91 16.11 16.29 16.47 16.66 16.84 17.01 17.18 17.34 17.48 17.61 17.72 17.84 17.99 18.13 18.28 18.44 18.6 18.75 18.89 19.01 19.28 31.54 Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Silty Sand to Sandy Silt Silty Sand to Sandy Silt Sand to Silty Sand Sand to Silty Sand Silty Sand to Sandy Silt Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Silty Sand to Sandy Silt Sandy Silt to Clayey Silt Sand to Silty Sand Silty Sand to Sandy Silt Silty Sand to Sandy Silt Sandy Silt to Clayey Silt Silty Sand to Sandy Silt Silty Sand to Sandy Silt Sandy Silt to Clayey Silt Sand to Silty Sand Silty Sand to Sandy Silt Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sandy Silt to Clayey Silt Sandy Silt to Clayey Silt Silty Clay to Clay Silty Clay to Clay Silty Clay to Clay Sandy Silt to Clayey Silt Sandy Silt to Clayey Silt Silty Clay to Clay Clay Silty Sand to Sandy Silt Silty Sand to Sandy Silt Sand to Silty Sand Silty Sand to Sandy Silt Sand to Clayey Sand Very Stiff Fine Grained Very Stiff Fine Grained Sandy Silt to Clayey Silt Silty Sand to Sandy Silt Silty Sand to Sandy Silt Sandy Silt to Clayey Silt Silty Sand to Sandy Silt Silty Sand to Sandy Silt Very Stiff Fine Grained Silty Sand to Sandy Silt Sensitive Fine Grained Sensitive Fine Grained 17 16 12 12 12 9 10 11 10 9 12 10 13 15 10 14 11 12 13 9 15 17 13 14 12 15 8 6 3 3 3 7 20 26 26 15 21 41 59 84 135 121 48 40 47 52 44 43 121 59 -33 -1 68.68 64.52 49.59 37.46 36.77 39.13 41.32 34.72 41.32 35.55 47.76 42.38 40.93 29.92 42.07 41.57 32.94 24.59 40.63 28.97 30.28 68.02 38.8 59.22 47.98 60.89 15.65 12.49 7.49 6.3 6.72 14.35 40.13 53.2 26.34 45.07 65.1 164.01 177.16 168.34 135.09 120.74 96.46 119.94 141.53 104.59 134.48 128.96 120.58 178.44 -32.8 -1.03 0.97 0.59 0.29 0.29 0.33 0.17 0.22 0.31 0.18 0.23 0.15 0.13 0.43 0.48 0.24 0.39 0.32 0.36 0.52 0.34 0.68 0.55 0.55 0.51 0.42 0.48 0.25 0.22 0.17 0.14 0.17 0.21 0.96 2.32 1.46 1.01 1.38 2.23 4.33 6.3 7.68 5.48 3.2 3.62 4.34 3.37 3.36 4.17 5.85 5.52 1.04 0.03 1.41 0.91 0.58 0.77 0.89 0.42 0.54 0.88 0.43 0.64 0.31 0.3 1.04 1.6 0.56 0.93 0.98 1.45 1.28 1.17 2.24 0.81 1.42 0.86 0.88 0.79 1.6 1.76 2.25 2.29 2.47 1.45 2.4 4.36 5.55 2.23 2.12 1.36 2.44 3.74 5.68 4.54 3.31 3.02 3.06 3.22 2.5 3.24 4.85 3.09 -3.16 -2.72

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Appendix B (Continued) Table B.2 CPT Sounding for NW CornerDepth (ft) 0 0.09 0.29 0.49 0.66 0.83 1.01 1.18 1.35 1.53 1.73 1.92 2.1 2.28 2.47 2.67 2.85 3.05 3.26 3.46 3.65 3.85 4.05 4.26 4.45 4.65 4.85 5.05 5.24 5.43 5.63 5.84 6.03 6.22 6.43 6.63 6.83 7.01 7.21 7.41 7.6 7.78 7.98 8.17 8.36 8.54 8.93 Soil Type No Reading Sand to Silty Sand Sand to Silty Sand Sand Sand Sand Sand Sand Sand Sand to Silty Sand Sand Sand Sand Sand Sand to Silty Sand Sand to Silty Sand Silty Sand to Sandy Silt Sandy Silt to Clayey Silt Clayey Silt to Silty Clay Clayey Silt to Silty Clay Silty Sand to Sandy Silt Silty Sand to Sandy Silt Silty Sand to Sandy Silt Silty Sand to Sandy Silt Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Silty Sand to Sandy Silt Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand N Value n/a 15 19 21 26 29 32 31 23 19 16 21 23 21 20 16 15 10 6 9 11 10 7 8 11 12 12 11 11 8 10 9 10 12 12 11 11 14 15 13 15 17 16 17 17 17 15 Tip Stress (bars) 20.2 59.61 77.42 108.45 131.43 148.8 159.68 154.82 118.69 77.48 81.09 107.67 119.02 105.34 83.14 67.3 45.65 19.76 12.93 18.82 35.08 30.86 22.89 24.25 43.76 47.76 49.09 46.51 45.1 34.91 30.41 37.02 39.93 48.12 48.2 44.43 46.4 58.36 61.25 55.45 62.8 68.82 64.22 71.4 69.07 69.32 59.75 Sleeve Stress (bars) 0.04 0.34 0.29 0.24 0.31 0.36 0.44 0.48 0.56 0.5 0.51 0.43 0.43 0.48 0.57 0.58 0.49 0.26 0.39 0.56 0.6 0.37 0.29 0.12 0.15 0.14 0.13 0.1 0.1 0.17 0.16 0.19 0.16 0.09 0.17 0.13 0.23 0.23 0.29 0.27 0.19 0.28 0.29 0.36 0.36 0.37 0.21 Friction Ratio (%) 0.2 0.58 0.38 0.23 0.23 0.24 0.28 0.31 0.47 0.64 0.63 0.4 0.36 0.45 0.68 0.86 1.07 1.32 3.04 2.98 1.7 1.2 1.27 0.5 0.34 0.29 0.27 0.21 0.22 0.48 0.53 0.5 0.4 0.19 0.34 0.29 0.49 0.4 0.47 0.49 0.3 0.41 0.46 0.5 0.52 0.53 0.35

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Appendix B (Continued) Table B.2 (Continued)9.12 9.31 9.5 9.7 9.89 10.09 10.28 10.49 10.67 10.86 11.05 11.25 11.44 11.63 11.82 12.02 12.21 12.39 12.59 12.78 12.97 13.16 13.36 13.56 13.74 13.92 14.11 14.31 14.49 14.68 14.88 15.08 15.27 15.46 15.66 15.85 16.03 16.19 16.37 16.55 16.7 16.84 16.96 17.08 17.19 17.29 17.37 29.27 29.27 29.27 29.27 Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Silty Sand to Sandy Silt Silty Sand to Sandy Silt Sand to Silty Sand Sand to Silty Sand Sandy Silt to Clayey Silt Sandy Silt to Clayey Silt Silty Sand to Sandy Silt Sand to Silty Sand Sand to Silty Sand Sand to Silty Sand Silty Sand to Sandy Silt Silty Sand to Sandy Silt Silty Sand to Sandy Silt Silty Sand to Sandy Silt Silty Sand to Sandy Silt Sand to Silty Sand Silty Sand to Sandy Silt Silty Sand to Sandy Silt Clayey Silt to Silty Clay Clay Clay Clay Clay Clayey Silt to Silty Clay Silty S