ECCS-TC10-03-528

1

PLASTIC ROTATION CAPACITY OF MR BEAM-TO-COLUMN JOINTS UNDER CYCLIC LOADING

Daniel Grecea, Aurel Stratan, Dan DubinăDepartment of Steel Structures and Structural Mechanics, The Politehnica University of Timisoara

Summary

The paper presents the rotation capacity of some beam-to-column joints subjected to cyclic loading. After the presentation of the rotation capacity definition, it is interesting to present how this characteristic of the joints is reflected in different national seismic design codes. The paper ends with an original proposal for the evaluation of the rotation capacity for beam-to-column welded joints established from some experimental tests. 1. Introduction Modern design of steel structures has introduced in the last years the real behaviour of joints, modifying the joint classification. This classification of joints has implemented the new definitions of semi-rigid and partial resistant joints.

Modern codes like Eurocode 3 [1] are defining 3 main characteristics to be taken into account for the design of steel structures. These characteristics are the moment resistance (Mu), the initial rotational stiffness (Sj,ini) and the rotation capacity (Φu).

In the last years, it was demonstrated by different authors [2], [3], [4], [5], [6], that, the first two characteristics Mu and Sj,ini of the beam-to-column joints are not influencing very strongly the structural behaviour of a building subjected to seismic actions. In fact, these two characteristics are influencing only the storey drifts and the second order effects P-∆, which can be very well controlled.

It is quite normal that the seismic response of a structure to be influenced by the beam-to-column joint ductility, characterised by the rotation capacity. 2. Definition of Rotation Capacity Rotation capacity characterises the ability of a plastified joint to rotate while maintaining its design moment resistance. Figure 1 illustrates this definition. The elastic rotation Φel is reached at the level of the elastic design moment resistance. Another rotation Φtr is necessary to get to the plastic design moment resistance Mpl,Rd. The plastic rotation Φpl is defined as the interval between the point where the real moment rotation curve reaches the level of Mpl,Rd and the point where the real curve reaches this level again. Consequently the available rotation capacity of a joint is defined as the difference:

ΦΦΦΦpl

ΦΦΦΦ

Mj

Mpl,Rd

Mj,u

ΦΦΦΦCd

Mel,Rd

ΦΦΦΦel ΦΦΦΦXd

ΦΦΦΦtr

Figure 1: Definition of rotation capacity for a joint

ECCS-TC10-03-528

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XdCdpl Φ−Φ=Φ (1) For the verification of sufficient rotation capacity of members a simplified procedure is given in Eurocode 3 [1]. Due to their web and flange slenderness sections can be subdivided into four classes that allow different methods of analysis. An explicit verification of sufficient rotation capacity is not necessary.

Usually the semi-rigid steel joints are partial resistant, which means that their resistance is less than that of the connected members, but as a matter of fact the highest moments often occur at the joints and not in the beam or columns. Therefore in systems with semi-rigid and/or partial resistant connections, plastic hinges will most probably form at the joints and not in the connected members. To allow rigid-plastic analysis the rotation capacity of the joints has to be checked, after Eurocode 3, Part 1.8 [7], paragraph 5.1.3: Rigid-plastic global analysis: (4) The rotation capacity of the joints shall be verified to be capable to accept the rotations of the joints resulting from the analysis. At present, Eurocode 3, Part 1.8 [7], paragraph 6.4: Rotation capacity gives only some basic principles for the verification of sufficient rotation capacity for a limited number of beam-to-column joints with a specific failure mode. No general methods of verification exist: (4) A beam-to-column joint in which the moment resistance of the joint Mj,Rd is governed by the resistance of the column web panel in shear, may be assumed to have adequate rotation capacity for plastic global analysis, provided that ε≤ 69td w .(5) In a welded beam-to-column joint in which the column web is stiffened in compression but unstiffened in tension, provided that the moment resistance is not governed by the shear resistance of the column web panel, see (4), the rotation capacity ΦCd may be assumed to be not less than the value given by:

bcCd hh025.0=Φ (3) where: hb is the depth of the beam; hc is the depth of the column. (6) An unstiffened welded beam-to-column joint designed in conformity with the provisions of this section, may be assumed to have a rotation capacity ΦCd of at least 0.015 radians. (7) A joint with a bolted connection with end-plates or angle flange cleats may be assumed to have sufficient rotation capacity for plastic analysis, provided that both of the following conditions are satisfied: a) the moment resistance of the joint is governed by the resistance of either:

- the column flange in bending; - the beam end-plate or tension flange cleat in bending.

b) the thickness t of either the column flange or the beam end-plate or tension flange cleat (not necessary the same component as in (a)) satisfies:

yub ffd36.0t ≤ (4) where: d is the nominal diameter of the bolts; fub is the ultimate tensile strength of the bolts; fy is the yield strength of the relevant basic component. (8) In cases not covered by (3) to (7), the rotation capacity may be determined by testing in accordance with EN 1990. Alternatively, appropriate calculation models may be used, provided that they are based on the results of tests in accordance with EN 1990.

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As there should be a common basis for the definition for the rotation capacity of members and joints, there should also be a common method for the verification of rotation capacity of members and joints.

A first step towards a common method of verification of rotation capacity is a classification of joints analogous to that of members. [8] and [9] refer to this point and give definitions for joint classes: Class 1 joints: Ductile joints.

A ductile joint is able to develop its plastic moment resistance and to exhibit a sufficiently large rotation capacity.

Class 2 joints: Joints of intermediate ductility.

A joint of intermediate ductility is able to develop its plastic moment resistance but exhibits only a limited rotation capacity once this resistance is reached.

Class 3 joints: Non ductile joints.

Premature failure (due to instability or to brittle failure of one of the joint components) occurs within the joint before the moment resistance based on a full plastic redistribution of the internal forces is reached.

As for members this classification and the corresponding deemed-to-satisfy criteria must be based on thorough scientific investigations. During this procedure a very strict distinction has to be made between the test level and the model level.

3. Beam-to-column joint ductility in seismic design codes for MR steel frames 3.1. Eurocode 8 (EC8-94) According to Eurocode 8 [10], connections in dissipative zones shall have sufficient overstrength to allow for yielding of the connected parts. For these overstrength verifications an appropriate estimation of actual value of the yield strength of the connected parts shall be made. Where more precise values are not available, the maximum value of the yield strength shall be used.

Non dissipative connections of dissipative members made by means of butt welds or full penetration welds are deemed to satisfy the overstrength criterion.

For fillet weld or bolted non dissipative connections, the following requirement should be met:

fyovd R10.1R γ≥ (2) where Rd is the resistance of the connection according to clause 6 of EN 1993-1-1 and Rfy is the plastic resistance of the connected dissipative member based on the design yield stress of material as defined in EN 1993-1-1. The effectiveness of such connection devices and their strength under cyclic loading shall be established by tests, to the satisfaction of the National Authorities. For bolted shear connections, the design shear resistance of the bolts should be higher than 1.2 times the design bearing resistance. Dissipative semi-rigid and/or partial strength connections are permitted, provided that all of the following conditions are satisfied: a) the connections have a rotation capacity consistent with the deformations; b) members framing into the connections are demonstrated to be stable at ultimate limit state (ULS); c) the effect of connections deformation on global drift is taken into account using non linear static (pushover) global analysis or non linear time history analysis.

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The overstrength condition for connections need not apply if the connections are designed in a manner enabling them to contribute significantly to the energy dissipation necessary to achieve the chosen q-factor. The connection design should be such that the plastic rotation capacity pθ in the plastic hinge is not less than 35 mrad for structures of ductility class H and 25 mrad for structures of ductility class M with q>2. 3.2. French code PS-92 According to the French code PS-92 [11] there is no specification concerning the rotation capacity of the beam-to-column joints.

Without a scientific specification established and validated by experience, using of semi-rigid connections is not authorised.

Connections made by means of full penetration welds are deemed to satisfy the overstrength criterion.

For fillet weld connections with partial or without penetration and for bolted connections the following requirement shall be met :

S Rd d E≤ / γ (5) where: Rd is the resistance of the connection γ E is the partial safety factor =1 for connections in non-dissipative zones =1.2 for connections in dissipative zones Sd is the design value of acting efforts (bending moment, shear force and axial force). Concerning bolted shear connections, only connections with pre-stressed high resistant bolts or connections with calibrated bolts are allowed to be used in dissipative zones. Tensioned bolted connections have to be used with pre-stressed high resistant bolts. 3.3. American code UBC-97 According to the UBC-97 [12], the purpose of the earthquake provisions herein is primarily to safeguard against major structural failures and loss of life, not to limit damage or maintain function.

Structures and portions thereof shall, as a minimum, be designed and constructed to resist the effects of seismic ground motions as provided in this division.

For this purpose, Moment Resisting Frames (MRF), defined as a frame in which members and joints are capable of resisting forces primarily by flexure, are divided as follows:

Ordinary Moment-Resisting Frame (OMRF) is a moment -resisting frame not meeting special detailing requirements for ductile behaviour.

Special Moment-Resisting Frame (SMRF) is a moment-resisting frame specially detailed to provide ductile behaviour and comply with the specific requirements.

Intermediate Moment-Resisting Frame (IMRF) is considered for concrete frames only, which are designed in accordance with specific requirements. UBC-97 code has no specifications concerning the rotation capacity of the beam-to-column joints. Anyhow, some interesting specifications are made for the first two categories of frames mentioned above.

All beam-to-column and column-to-beam connections in OMRF which resist seismic forces shall meet one of the following requirements: a. For FR (fully restrained) connections, the required flexural strength, Mu, of a column-to-beam

joint is not required to exceed the nominal plastic flexural strength of the connection.

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b. Either FR and PR (partially restrained) connections shall meet the following : 1. The connections have been demonstrated by cyclic tests to have adequate rotation capacity at

a story drift calculated as a horizontal load of 0.4R × E, (where the term 0.4R is equal to or greater than 1.0 and where E is the earthquake load and R is the response modification factor).

2. The additional drift due to PR connections shall be considered in design. FR and PR connections are described in detail in Sect. A2 of the Specification. For SMRF, the beam-to-column joints have to satisfy that the required flexural strength, Mu,

of each beam-to-column joint shall be lesser of the following quantities: 1. The plastic bending moment, Mp, of the beam. 2. The moment resulting from the panel zone nominal shear strength, Vn.

No other precisely specification concerning the rotation capacity of the joints is made by the UBC code. 3.4. American code AISC – 2002 draft According to AISC-97 [13], Moment Frame (MF) defined as a building frame system in which seismic shear forces are resisted by shear and flexure in members and connections of the frame are divided as follows:

- Special Moment Frame (SMF) - Intermediate Moment Frame (IMF) - Ordinary Moment Frame (OMF)

Special Moment Frames (SMF) are expected to withstand significant inelastic deformations when subjected to the forces resulting from the motions of the Design Earthquake. SMF shall meet the following requirements: • The connection must be capable of sustaining an Interstory Drift Angle of at least 0.04 radians. • The flexural strength of the connection, determined at the column face, must equal at least 80

percent of the nominal plastic moment of the connected beam at an Interstory Drift Angle of 0.04 radians.

These requirements could be demonstrated by one of the following: • Use a connection Prequalified for SMF. • Provide qualifying cyclic test results. Results of at least two cyclic connection tests shall be

provided and are permitted to be based on one of the following: a. Tests reported in research literature or documented tests performed for other projects that are demonstrated to reasonably match project conditions. b. Tests that are conducted specifically for the project and are representative of project member sizes, material strengths, connection configurations, and matching connection processes.

Intermediate Moment Frames (IMF) are expected to withstand limited inelastic deformations in their members and connections when subjected to the forces resulting from the motions of the Design Earthquake. All beam-to-column joints and connections used in the Seismic Load Resisting System shall satisfy the following three requirements: • The connection must be capable of sustaining an Interstory Drift Angle of at least 0.02 radians. • The flexural strength of the connection, determined at the column face, must equal at least 80

percent of the nominal plastic moment of the connected beam at an Interstory Drift Angle of 0.02 radians.

These requirements shall be demonstrated by one of the following: • Use a connection prequalified for IMF. • Provide qualifying cyclic test results. Results of at least two non-identical cyclic connection

tests shall be provided and are permitted to be based on one of the following:

ECCS-TC10-03-528

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a. Tests reported in research literature or documented tests performed for other projects that are demonstrated to reasonably match project conditions. b. Tests that are conducted specifically for the project and are representative of project member sizes, material strengths, connection configurations, and matching connection processes.

Ordinary Moment Frames (OMF) are expected to withstand minimal inelastic deformations in their members and connections when subjected to the forces resulting from the motions of the Design Earthquake. OMF shall meet the following requirements. Beam-to-column connections shall be made with welds and/or high-strength bolts. Connections are permitted to be FR or PR moment connections as follows: • FR moment connections that are part of the Seismic Load Resisting System shall be designed

for a required flexural strength Mu that is at least equal to 1.1RyMp of the beam or girder or the maximum moment that can be delivered by the system, whichever is less.

• PR moment connections are permitted when the following requirements are met: a. Such connections shall provide for the design strength as specified above b. The nominal flexural strength of the connection shall be no less than 50 percent of Mp of the connected beam or column, whichever is less. c. The stiffness and strength of the PR moment connections shall be considered in the design, including the effect on overall frame stability.

Even if the rotation capacity of the beam-to-column joints is connected with the classification of frames, the AISC code is not providing any formula for the evaluation of this very important characteristic. Special attention was paid to the both American seismic design codes UBC-97 and AISC-2002, because these two codes are the most recent and they are integrating the conclusions post-Northridge earthquake (1994). These conclusions were very interesting for steel structures because Northridge earthquake affected especially the joints of the steel structures. 3.5. Japanese code AIJLSD - 90 The Japanese code AIJLSD-90 [14] provides specifications concerning only the frame classification, according to the classification of members and member cross-sections, storey-drift and overstrength of joints in comparison with the connected members. Any specification concerning the rotation capacity of beam-to-column joints is not made. 3.6. Romanian code P100 - 92 Romanian code P100 – 92 [15] is making a classification of structures similar with Eurocode 8, with storey-drift specifications and overstrength conditions concerning the structural joints, but with no specific provisions about the rotation capacity of the beam-to-column joints. 4. Tests and estimation of rotation capacity values for welded joints (INSA Rennes) Aribert & Grecea [5] have developed an experimental research program at INSA Rennes. This research program dealt with 8 beam-to-column welded joints of different sizes under monotonic and repeated cyclic loading (Table 1). The specimens were major axis joints with a symmetrical cruciform arrangement comprising a H or I column connected to two cantilever beams by full penetration butt welds with double bevel in the beam flanges. No transverse stiffener was welded in the compression zone of the column web, so that the static partial resistance of the joints was governed by local buckling of the column web.

ECCS-TC10-03-528

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The tests were performed according to the Recommended Testing Procedure of ECCS (1985). Each type of joint was subject first of all to a monotonic loading (CPP11, CPP13, CPP15 and CPP17), which allowed to determine the main static characteristics as the moment resistance, the initial rotational stiffness, the maximum elastic rotation and the rotation capacity.

After the determination of these characteristics, the elements were subject to cyclic reversal loading (CPP12, CPP14, CPP16 and CPP18). The cyclic moment-rotation curves for the four tested joints are presented in Figure 2.

The cyclic behaviour of this type of joint may be characterised by a good regularity in the loops shape with more or less deterioration between loops. This regularity could be explained by the sufficient continuity of internal forces and the reasonable value of the local buckling slenderness of the column web subject to transverse compression (non-dimensional slenderness λ ≤ 0 8. ). Generally the failure mode occurred by local buckling of the column web. Nevertheless, due to plastic deformations developed in the column web under alternate compression and tension, a crack in the web panel near the flange was observed in a few cases. Simultaneously, in the column flange subject to alternate bending, another crack started close to the weld connecting the column flange with the beam flange. Experimental values of ultimate resistance moment and rotation capacity obtained in both monotonic and cyclic reversal loading can be compared in Table 2.

From the moment-rotation curves, it is observed that the ultimate moment and the initial stiffness of the joints are not strongly influenced by the repeated cyclic loading, so that in seismic design the corresponding formulae given in Eurocode 3 for the case of static loading can be used, as reasonable approximations.

On the other hand it appears clearly that the rotation capacity of the joints is systematically reduced by a factor about 2. As in the literature there are no formulae to evaluate the rotation capacity of the relevant joints, the present authors have proposed the following one for monotonic loading:

Φ u b ch h= 0 030. (6)

and the following one for cyclic loading :

Φ u b ch h= 0 015. (7)

where hb is the beam depth and hc the column depth (it should be noted that φu would be proportional to ratio hb/hc, and not to ratio hc/hb as mentioned in Clause J.5(5) of Annex J of EC3).

In addition the failure mode may be different under repeated cyclic loading in comparison with the static one; for example fracture of the column flange and web may occur instead of local buckling of the column web, which can be controlled only by design methods including low-cycle fatigue phenomena and damage models. Table 1 Experimental elements Test specimens Column Beam Loading type Failure mode CPP 11 HEB 200 IPE 360 Monotonic Buckling of column web CPP 12 HEB 200 IPE 360 Cyclic Reversal Buckling of column web with

fracture of column flange and web CPP 13 IPE 360 IPE 360 Monotonic Buckling of column web CPP 14 IPE 360 IPE 360 Cyclic Reversal Buckling of column web CPP 15 HEB 300 IPE 360 Monotonic Buckling of column web CPP 16 HEB 300 IPE 360 Cyclic Reversal Buckling of column web CPP 17 HEB 300 IPE 450 Monotonic Buckling of column web CPP 18 HEB 300 IPE 450 Cyclic Reversal Buckling of column web with

fracture of column flange and web

ECCS-TC10-03-528

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Test CPP 12

-300

-200

-100

0

100

200

300

-0,03 -0,02 -0,01 0 0,01 0,02 0,03 0,04

ΦΦΦΦ [rad]

M [kNm]Test CPP14

-200

-150

-100

-50

0

50

100

150

200

-0,04 -0,03 -0,02 -0,01 0 0,01 0,02 0,03 0,04

ΦΦΦΦ [rad]

M [kNm]

Test CPP16

-400

-300

-200

-100

0

100

200

300

400

-0,02 -0,01 0 0,01 0,02 0,03

ΦΦΦΦ [rad]

M [kNm]Test CPP18

-600

-400

-200

0

200

400

600

-0,05 -0,04 -0,03 -0,02 -0,01 0 0,01 0,02

ΦΦΦΦ [rad]

M [kNm]

Figure 2: Cyclic behaviour of the joints

Table 2 Comparison between joint characteristics under cyclic and monotonic loadings CPP11 CPP12 CPP13 CPP14 CPP15 CPP16 CPP17 CPP18 Mu [kNm] 230.0 253.0 166.9 180.3 349.2 368.5 467.5 486.2 Φu [rad] 0.064 0.031 0.045 0.023 0.045 0.020 0.052 0.030

5. Tests and estimation of rotation capacity values for welded and bolted joints (U.P. Timisoara)

The present work describes investigations on beam-to-column joints, carried out at the laboratory of steel structures at the Civil Engineering Faculty of Timisoara. First are described the joints that have been tested, their characteristics, the loading system and procedure. Further are given the theoretical characteristics of the joints, computed according to EUROCODE 3. The results of the tests present the experimental characteristics of the tested joints and the behavioural curves. Finally, a comparison between the theoretical (by EUROCODE 3) and the experimental characteristics is made, as well as the resulting conclusions. 5.1. Specimens and testing set-up description Three typologies of beam-to-column joints have been tested from a total of 12 specimens. All the joints are double sided. For all the specimens the design steel grade was S235 (fy=235 N/mm2,fu=360 N/mm2), beams being IPE 360 and columns HEB 300 (Figure 3).

ECCS-TC10-03-528

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1100 1100

Column HEB 300Beam IPE 360

Connection

symmetrical loadingApplied force

(a) supplementary support

Column HEB 300

11001100

Connection

anti-symmetrical loading

Beam IPE 360

Applied force

(b) Figure 3: Static scheme and general description of the specimens: (a) symmetrical loading and (b)

anti-symmetrical loading

Actuator 1000kN

Supports

2550

725

1225

2200

Specimen

Figure 4: Testing set-up for symmetrical loading: scheme and real set-up Two types of loading were applied: symmetrical and anti-symmetrical (Figure 4 and Figure 5) and three connection typologies were tested (Figure 6): Type 1 – Figure 6 a:

- 2 symmetric cruciform extended end plate bolted connections (prestressed 10.9 M20 bolts) – specimens XS-EP1 and XS-EP2

- 2 anti-symmetric cruciform extended end plate bolted connections (prestressed 10.9 M20 bolts) specimens - XU-EP1 and XU-EP2

Type 2 – Figure 6 b: - 2 symmetric cruciform welded connections (full-penetration welds) specimens - XS-W1

and XS-W2 - 2 anti-symmetric cruciform welded connections (full-penetration welds)

specimens - XU-W1 and XU-W2 Type 3 – Figure 6 c:

- 2 symmetric cruciform connections with welded cover plates (full-penetration welds) and welded web plate (bolted for erection) specimens - XS-CWP1 and XS-CWP2

- 2 anti-symmetric cruciform connections with welded cover plates (full-penetration welds) and welded web plate (bolted for erection) specimens - XU-CWP1 and XU-CWP2

ECCS-TC10-03-528

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SupportsSpecimen

Actuator 1000kN

2200

1100

1155

Figure 5: Testing set-up for anti-symmetrical loading: scheme and real set-up

10M20 gr 10.9

Column HEB 300

Beam IPE360

End Plate t=20

3M20 gr6.6Equal strength weld

Column HEB 300

Beam IPE360

Equal Strenght weld

Column HEB 300

Supplementary web plate

Beam IPE360

Welded cover plate

(a) (b) (c) Figure 6: Connection configurations: (a) bolted, (b) welded and (c) with cover welded plate

The joints (three different joint configurations and two load types) have been designed according to Eurocode 3, Annex J. The joints’ classification, according to the above mentioned code is given in Table 3. Table 3 Joints’ stiffness and resistance classification according to EC3

SPECIMEN EC3 Stiffness Classification

EC3 Resistance Classification

Weakest Component

XS-EP Semi-rigid Partial-resistant End-plate in bending XS-W Rigid Equal-resistant Beam fl. & web in compr.XS-CWP Rigid Full-resistant Beam fl. & web in compr.XU-EP Semi-rigid Partial-resistant End-plate in bending XU-W Semi-rigid Partial-resistant Web panel in shear XU-CWP Semi-rigid Partial-resistant Web panel in shear

5.2. Loading history The loading history was made according to the ECCS Recommendations simplified procedure, (Figure 7), in which were performed three cycles for each even multiplier of the displacement ey,which represents the characteristic conventional yielding displacement of the joint. It was assumed as the displacement at the column end (for both symmetrical and anti-symmetrical). Prior the plastic cycles, the simplified ECCS procedure was used in order to find the displacement ey and the corresponding force Fy, so as to ensure that at least four levels of displacement have been performed before the conventional yielding displacement.

ECCS-TC10-03-528

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-8

-6

-4

-2

0

2

4

6

8 e/ey

time

Figure 7: Load history - recommended ECCS procedure

The end of the experiment was considered when, the final force load applied to the joint was at the most half of the maximum load applied to that joint. In some cases, due to premature failure of joints (XS-W1), or due to unexpected events during the test (one support felt during the XU-W2 testing), the experiment was stopped earlier. The applied loading speed was quasi-static. The total duration of a cycle was 8 minutes, the loading speed depending on the amplitude of displacement imposed. This slow rate of loading was imposed actually by the data acquisition system rate of recording. 5.3. Testing results XS-EP specimens – Symmetrical Joints

XS-EP2

-400

-300

-200

-100

0

100

200

300

400

-0.06 -0.04 -0.02 0 0.02 0.04 0.06

TOTAL JOINT ROTATION [Rad]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

(a)

XS-EP2

-400

-300

-200

-100

0

100

200

300

400

-0.06 -0.04 -0.02 0 0.02 0.04 0.06

TOTAL JOINT ROTATION [rad.]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

(b) Figure 8: Cyclic Moment-Rotation curve (a) and the envelope curve (b) for specimen XS-EP2

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Figure 9: Failure of specimen XS-EP2 XS-W specimens – Symmetrical Joints

XS-W2

-500

-400

-300

-200

-100

0

100

200

300

400

500

-0.015 -0.01 -0.005 0 0.005 0.01 0.015 0.02

TOTAL JOINT ROTATION [rad.]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

(a)

XS-W2

-500

-400

-300

-200

-100

0

100

200

300

400

500

-0.015 -0.01 -0.005 0 0.005 0.01 0.015 0.02

TOTAL JOINT ROTATION [rad.]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

(b) Figure 10: Cyclic Moment-Rotation curve (a) and the envelope curve (b) for specimen XS-W2

Figure 11: Failure of specimen XS-W2

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XS-CWP specimens – Symmetrical Joints

XS-CWP1

-600

-400

-200

0

200

400

600

-0.04 -0.02 0 0.02 0.04 0.06

TOTAL JOIT ROTATION [rad.]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

(a)

XS-CWP1

-600

-400

-200

0

200

400

600

-0.04 -0.02 0 0.02 0.04

TOTAL JOINT ROTATION [rad.]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

(b) Figure 12: Cyclic Moment-Rotation curve (a) and the envelope curve (b) for specimen XS-CWP1

Figure 13: Failure of specimen XS-CWP1 XU-EP specimens – Anti-Symmetrical Joints

XU-EP1

-350

-250

-150

-50

50

150

250

350

-0.06 -0.04 -0.02 0 0.02 0.04 0.06

TOTAL JOINT ROTATION [rad]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

XU-EP1

-350

-250

-150

-50

50

150

250

350

-0.08 -0.06 -0.04 -0.02 0 0.02 0.04 0.06 0.08

TOTAL JOINT ROTATION [rad]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

a) Cyclic behaviour b) Envelope of hysteresis loops

Figure 14: Total joint rotation versus moment relationships for XU-EP1 specimen

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a) XU-EP1 b) XU-EP2 Figure 15: Rupture of end plate (a) and failure of beam near the beam to end plate connection (b)

XU-W specimens – Anti-Symmetrical Joints

XU-W1

-350

-250

-150

-50

50

150

250

350

-0.06 -0.04 -0.02 0 0.02 0.04 0.06

TOTAL JOINT ROTATION [rad]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

XU-W1

-350

-250

-150

-50

50

150

250

350

-0.06 -0.04 -0.02 0 0.02 0.04 0.06

TOTAL JOINT ROTATION [rad]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

a) Cyclic behaviour b) Envelope of hysteresis loops Figure 16: Total joint rotation versus moment relationships for XU-W1 specimen

a) XU-W1 b) XU-W1 Figure 17: Rupture of column web (a) and failure of beam top flange (b)

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XU-CWP specimens – Anti-Symmetrical Joints

XU-CWP1

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-150

-50

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350

-0.07 -0.05 -0.03 -0.01 0.01 0.03 0.05 0.07

TOTAL JOINT ROTATION [rad]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

XU-CWP1

-350

-250

-150

-50

50

150

250

350

-0.08 -0.06 -0.04 -0.02 0 0.02 0.04 0.06 0.08

TOTAL JOINT ROTATION [rad]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

a) Cyclic behaviour b) Envelope of hysteresis loops Figure 18: Total joint rotation versus moment relationships for XU-CWP1 specimen

a) XU-CWP1 b) XU-CWP2 Figure 19: Failure of column web (a) and crack through the column flange and web (b)

General Results from the Tests Table 4 comprises the main parameters monitored during the tests: Py – the force corresponding to joint yielding δy,– the yielding displacement corresponding to PyMmax – the maximul bending moment obtained at the column face for the test φu – the ultimate (maximum) rotation of the joint Monitored parameters are also the maximum energy dissipated in a cycle and the total energy dissipated during a test. The number of plastic cycles to failure is considered an important parameter describing the plastic performances of the joint. Table 4: Main characteristics of cyclic tests

SPECIMEN Py[kN]

δy[mm]

maxM[kNm]

φu+

[rad] φu

-

[rad] Max. En/cycl.

[kNm rad] Total En. [kNm rad]

Nr. of pl. cycles

XS-EP 1 569.12 10 334.17 0.031 0.033 20.83 76.74 11

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XS-EP 2 522.38 6.5 337.94 0.039 0.037 20.84 120.15 16XS-W 1 642.40 4.36 437.70 0.028 0.010 19.71 125.20 16XS-W 2 677.88 4.85 412.12 0.017 0.013 18.05 64.69 10XS-CWP 1 678.40 5.60 542.01 0.036 0.038 45.06 390.00 21XS-CWP 2 Accidental failure of the column loading end plate XU-EP 1 170.0 15.00 263.7 0.055 0.060 39.4 661.50 34XU-EP 2 170.0 15.00 256.3 0.057 0.062 42.1 924.60 37XU-W 1 177.0 12.80 247.8 0.052 0.051 39.3 721.00 30XU-W 2 177.0 12.80 252.1 0.052 0.050 38.0 611.70 23XU-CWP 1 178.3 11.00 287.2 0.054 0.064 45.7 1666.80 50XU-CWP 2 178.3 11.00 301.5 0.060 0.060 52.0 1051.20 31

It can be observed that generally, there are not big differences between the two specimens of the same type. Anyway, a few comments should be added. First, the XS-W1 specimen was deliberately different from the XS-W2 specimen, by re-welding of the weld roots of the former specimen (as explained earlier). Secondly, although the results are similar in terms of maximum values of rotations and moments resisted by the joints, the failure mechanism and the number of plastic cycles are sometimes substantially different. This is particularly true for the XU-CWP joints. Due to the different statical schemes for the two types of loading, the bending moment in the node for anti-symmetrical scheme is obtained by half the force needed for the symmetrical one. But this is not the cause of the drastic drop in the yield force Py from the XS to XU series. A change in the loading type caused also a change of the joint resistive components. Test results are in accordance with the expected joint behaviour as it can be seen in Table 3. In what concerns the XS series, joint resistance and rigidity are expected to increase in the range EP-W-CWP. Results in terms of maximum moment attained during testing confirm this trend. It should be noted that the maximum moment for all joints is computed at the column face. The behaviour of the three types of joints is quite different: • XS-EP joints showed a good ductility mainly due to the end plate in bending and partially due

to local buckling of beam flanges. Anyway, failure was achieved not only by rupture of the end plate, but also by rupture of the beam flange at its connection to the end plate (in the weld or in the heat affected zone). Mean maximum rotation attained (0.035 radians) shows a good plastic behaviour of XS-EP joints. The failure was a ductile one.

• Ductility of XS-W joints was mainly affected by the brittle and sudden failure of the beam to column connection. The mean maximum rotation (0.017 radians) proves this fact. This type of joint is especially affected by the quality of welds at the beam to column connection.

• The objective of XS-CWP joints, reinforced at the beam to column connection, was accomplished, the plastic zone shifting from the column face connection into the beam. Therefore, its ductility is practically given by the beam ductility. The equivalent rotation at the column face (0.037 radians) is smaller than the real one in the plastic hinge.

The dissipated energy, both the maximum and the cumulated one have comparable values for XS-EP and XS-W specimens, and are substantially greater for XS-CWP specimen. The number of hysteresis loops to failure is again much greater for the latter case. In the case of XU series, the maximum moment attained is expected to have close values, taking into consideration that the main resisting component is the web panel in shear. This was proved to be true, with the specification that the XU-W joint showed the smallest maximum moments and the XU-CWP the biggest values. Their ductility is comparable (0.051-0.059 radians mean values), all the joints proving good ductility. Behaviour of the three types of joints is in general similar, being governed by the shear behaviour of the column web panel, but there are also some particularities:

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17

• Beside shear in the panel zone, behaviour of XU-EP joints was influenced by the end plate in bending. At the first plastic cycles the ductility demand was distributed between the two components, participating together to the plastic excursions. While panel zone had stable hysteresis loops, the behaviour of the end plate was characterised by significant degradation (due to loosening of bolts and rupture of the extended part of the end plate).

• In the case of XU-W joints, plasticity was spread between the panel zone and in a smaller extent the beam flanges. These joints have been the least ductile, failure occurring by brittle fracture of beam flanges and pullout of the column flange.

• The web panel governed exclusively the behaviour of XU-CWP joints. Failure was due to ductile degradation of the panel zone, which was finally torn apart. A concern should be expressed here about this type of joints, as the second specimen failed by complete rupture of the column flange.

The maximum dissipated energy is higher for the XU-CWP joints. The cumulated energy is considerably higher for the same type of joints. This is partially caused by the increased number of plastic cycles. In what concerns the differences between the XS and XU series, change of loading type led to important differences between the two series. Generally, a drop in maximum moment is observed for the anti-symmetrical loading. Anyway, this drop is different among the connection types as follows: 15% for the end plate joints and about 40% in the case of welded and cover plated joints. This fact is explained by close resistance of the extended end plate and the web panel, both components being involved in the plastic mechanism. While for the other two cases the web panel was the main participating component. Joint rotations are considerably higher for XU series. Improved ductility in the case of XU joints is given by good rotation capacity and stable hysteresis loops of the web panel in shear. Anti-symmetrical joints have generally increased energy dissipation capacity with respect to the symmetrical ones. This fact is given by the increase of both maximum energy dissipated per cycle and number of cycles (case of EP and W joints), or only increased number of cycles (case of CWP joints). 5.4. Comparison between the EC3 and Experimental Results Table 8 comprises the results of the experimental tests compared to that of EC 3 – Annex J, in terms of joint bending moments, rotational stiffness and ultimate rotation attained. It should be noted that for this comparison, the joint characteristics are computed with the measured strengths and dimensions of the joint components. Table 8 Comparison between computed and experimental joint characteristics

SPECIMEN (exp),RdjM

[kNm]

)(,thRdjM

[kNm]

(exp),inijS

[kNm/rad]

)(,thinijS

[kNm/rad]

(exp)yφ

[rad]

(exp)uφ

[rad] XS-EP 1 255.6 262.7 69539 142932.2 0.0038 0.033XS-EP 2 288.9 261.3 44205 140886.8 0.0063 0.038XS-W 1 305.6 309.3 333953 ∞ 0.0009 0.029XS-W 2 277.8 317.6 321569 ∞ 0.0009 0.016XS-CWP 1 316.7 452.1 366309 ∞ 0.0009 0.038XS-CWP 2 ** 449.0 ** ∞ ** **XU-EP 1 146.7 169.2 44081 43727.2 0.0033 0.060XU-EP 2 157.8 169.1 49004 43718.2 0.0028 0.062XU-W 1 113.3 163.6 63102 68792.1 0.0020 0.052XU-W 2 131.1 164.1 49681 69062.1 0.0026 0.052XU-CWP 1 131.1 178.6 60712 75597.2 0.0022 0.064

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XU-CWP 2 164.4 177.4 58453 74963.1 0.0026 0.060

The yielding bending moment (exp)RdM is computed according to the ECCS procedure, as in Figure

20, resulting at the intersection of the Sj,ini line and the tangent to the envelope curve Sj,ini/10 line. The intersection point corresponds to the pair ( (exp)

RdM ,φy). Comparing the experimental and computed values of joint moment capacity, it can be observed that generally, close values are obtained for the XS series. An exception is the XS-CWP joint, which showed considerably lower experimental value. In the case of XU series, all experimental values are lower than the ones computed by Annex J of EC3. The difference between the computed and measured yielding bending moment could be explained by several causes: • Annex J of EC3 does not consider cyclic loading neither strain hardening • On the other hand, the procedure applied for determining the experimental yielding moments is

a conventional one and is greatly influenced by the initial stiffness of the joint

XS-EP2

0

50

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150

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250

300

350

400

0 0.005 0.01 0.015 0.02 0.025 0.03 0.035 0.04 0.045

JOINT ROTATION [RAD]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

Sj,ini/10

Sj,ini

Mj,Rd

FFFFy

Figure 20: Definition of (exp)RdM and φy.

In what concerns the initial stiffness of the joints, numerical and experimental results agree fairly well for the XU series, while significant differences are noticed for XS series. Anyway, stiffness is much lower for the anti-symmetrical joints both from experimental and computed stiffness values. This fact is again given by the deformability of the panel zone. 5.5. Concluding remarks As it was expected, the loading type (symmetrical or anti-symmetrical) significantly affects the response parameters of beam-to-column joints. The main component that brings the difference is the panel zone in shear. The most important consequences on the cyclic behaviour of beam-column joints are the reduced moment capacity and, (in general) increased ductility with more stable hysteresis loops in the case of anti-symmetrical loading. These tests were conducted under limiting cases of load asymmetry. The two loading types affect significantly joint properties in terms of initial stiffness, moment and rotation capacities. Therefore, when modelling the joint for structural analysis, different characteristics should be used for gravitational and lateral loading.

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Investigation of the different joint typologies revealed the importance of detailing of the connection and the welding procedure, as well as its quality. Defective welding was responsible for such phenomena as crack initiation and early cracks in welds or heat-affected zone. Bolted end-plate joints showed an increased rotation capacity and more ductile behaviour with respect to welded joints. Extended end plate connections should be designed so as to prevent brittle failure by bolt rupture. Loosening of bolts during cycle reversals has lead to stiffness degradation. Another aspect characteristic to anti-symmetrical bolted joints is the distribution of ductility demands between the end-plate (connection) and the panel zone. Generally, failure was brittle for welded joints and ductile for the other ones in the case of symmetrical loading. The ductile behaviour was due to connection (bolted joint) and due to shifting of the plastic hinge away from the column face in the case of CWP joint. Participation of panel zone to the plastic mechanism significantly increased the ductility of the anti-symmetrically-loaded joints. Anyway, welded joints failed in a brittle manner in this case, too. Generally, the joint with cover and web plates showed a good behaviour. Anyway, care should be taken when designing such joints due to potential problems caused by increased moment at the column face.

6. Cyclic Tests on Bolted Steel and Composite Double-Sided Beam-to-Column Joints (U.P. Timisoara)

An alternative to the “standard” European column cross-section (hot-rolled I profiles) is the use of X-shaped cross-sections, built-up of two hot-rolled profiles welded along the median axis or built-up sections made out from welded plates, as shown in Figure 21.

7Ø12 PC52 7Ø12 PC522Ø10 PC52

2Ø12 PC52

2Ø12 PC52

3Ø10 PC52

8M20 gr.10.9

290

20 360 20

14

12

Beam

170

Column 8 top & bot. fl.

5 .

12

6M20 gr.10.9

Column29

0

5 .

8 top & bot. fl.

Beam

120

600

R20

r.c. slab

R20

(a) (b) Figure 21: Connection configurations for BX-S series (a) and BX-C series (b) of joints.

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6.1. Experimental tests on Bare-steel joints The testing program comprised six specimens: three joints under symmetrical loading (BX-SS, see Figure 22a), and three joints under anti-symmetrical loading (BX-SU, see Figure 22b). The bolts have been fully preloaded, except for the last joint of each series, which have been preloaded to 50% only. Tests were performed in accordance with the ECCS Recommendations complete procedure (ECCS 1985). The first specimen from each series was tested monotonically, in order to determine the conventional yield displacement ey. Specimen failure was considered at 50% reduction of the maximum load applied during the loading history. The load was applied quasi-statically, under displacement control.

Actuator 1000kN

SupportsSpecimenSupports

SpecimenActuator 1000kN

(a) (b) Figure 22: Testing set-up for symmetrical loading (a) and for anti-symmetrical loading (b).

Table 9 and Figures 23, and 24 synthetically present the main experimental results expressed in terms of maximum plastic rotation maxϕϕϕϕ , maximum moment attained maxM , and the cumulated energy dissipated by the specimens during the entire loading history Etot.

Table 9 Experimental results for bare-steel joints. Specimen +

maxϕ , rad −maxϕ , rad +

maxM , kNm −maxM , kNm Etot, kNm rad

BX-SS-M 0.043 - 263.3 - 9.0 BX-SS-C1 0.028 0.021 271.6 259.1 41.5 BX-SS-C2 0.017 0.018 261.8 259.8 23.4 BX-SU-M 0.106 - 258.4 - 24.0 BX-SU-C1 0.073 0.055 269.4 240.6 135.8 BX-SU-C2 0.039 0.047 240.1 236.6 88.4

BX-SS-C1

-400

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-200

-100

0

100

200

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400

-0.1 -0.08 -0.06 -0.04 -0.02 0 0.02 0.04 0.06 0.08 0.1

TOTAL JOINT ROTATION [rad]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

(a)

BX-SU-C1

-400

-300

-200

-100

0

100

200

300

400

-0.1 -0.08 -0.06 -0.04 -0.02 0 0.02 0.04 0.06 0.08 0.1

TOTAL JOINT ROTATION [rad]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

(b) Figure 23: Moment - rotation relationships for cyclic specimens of the BX-S series

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(a) (b) Figure 24: BX-SS-C2: Weld failure (a); BX-SU-C1: End plate failure (b)

6.2. Experimental tests on Composite joints The steel part of the composite joints is similar to the bare-steel joints, except the end-plate, which is extended only at the bottom part (Figure 21), following the rationale that the slab reinforcement will compensate the missing bolt row at the top. The r.c. slab has a total depth of 120 mm, lying on a LINDAB LTP45x0.7 corrugated sheet. The slab width was taken equal to the effective width of the slab, computed according to Eurocode 4 (1992). Member dimensions and effective slab width have been determined by designing a three-bay (3x4.5m), three-storey (3x3.5m) moment resisting frame. The effective width of the slab is relatively small (600 mm), this needs to be recognised when considering the conclusions of the present study. Nelson connectors have been used, to achieve a full shear connection between the beam and the slab. The experimental program comprised three symmetrically loaded specimens (BX-CS series: one tested monotonically to negative moments – BX-CS-M1, one tested monotonically to positive moments – BX-CS-M2, and the third one tested under cyclic loading – BX-CS-C1) and three anti-symmetrical loading (BX-CU series: one tested monotonically – BX-CU-M, and the other two cyclically – BX-CU-C1 and BX-CU-C2). In the case of the BX-CU-C2 specimen, the ECCS procedure was not applied; instead, constant cycles of 6ey were used. In the case of composite joints all the bolts have been preloaded to 100% of the full preloading value. The results of tests on composite joints are summarised in Tables 10, and Figures 25 and 26. It is to be mentioned that in the case of anti-symmetrically loaded joints, the moments on the opposite sides of the column are not equal, due to different moment capacities and connection stiffness on the two sides. Only actuator force was recorded during the tests, so that it is not possible to determine directly the moments on the two sides of the column. Therefore, a mean value of the positive and negative moments was considered in the present study. Table 10 Experimental results for composite joints.

SPECIMEN +maxϕ , RAD −

maxϕ , RAD +maxM , KNM −

maxM , KNM ETOT, KNM RAD

BX-CS-M1 - 0.087 - 244.3 16.6 BX-CS-M2 0.033 - 316.1 - 8.4 BX-CS-C1 0.045 0.014 305.5 196.9 32.0 BX-CU-M* 0.090 198.2 15.2 BX-CU-C1* 0.058 190.2 104.5 BX-CU-C2* 0.032 200.9 32.2

* Mean values

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BX-CS-C1

-350

-250

-150

-50

50

150

250

350

-0.1 -0.08 -0.06 -0.04 -0.02 0 0.02 0.04 0.06 0.08 0.1

TOTAL JOINT ROTATION [rad]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

(a)

BX-CU-C1

-350

-250

-150

-50

50

150

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350

-0.1 -0.08 -0.06 -0.04 -0.02 0 0.02 0.04 0.06 0.08 0.1

TOTAL JOINT ROTATION [rad]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

(b) Figure 25: Moment - rotation relationships for cyclic specimens of the BX-C series

(a) (b) Figure 26: Failure of cyclically tested specimens form the BX-C series:

BX-CS-C1 (a) and BX-CU-C1 (b) 6.3. Comparison of test results to code results (Annex J of EC3 and EC4) Table 11 and Table 12 present the joint characteristics obtained from the tests and computed analytically in accordance to the Annex J of Eurocode 3 and Eurocode 4, respectively. For the analytical results, the real strengths and the measured dimensions of the members have been used. Steel joints The Annex J of Eurocode 3 gives the possibility of computing joints having I or H cross-sections (hot rolled or built-up profiles). The main difference between the behaviour of the I shaped profiles and the X-shaped ones is the increase in the shear area of the panel zone in the latter case, due to the presence of the supplementary flanges of the column. The X-shaped cross-sectional columns can be considered in Annex J by increasing accordingly the shear area of the panel zone. On the other hand, the Annex J specifications states that the shear area given by the panel zone can be supplemented by means of supplementary web plates, welded on one or both sides of the column web. In this way the shear area can be increased by a maximum area of bstwc (bs - web plate width, twc – the column web thickness) from one or two supplementary web plates (Figure 27b). In the case of X-shaped columns, due to the presence of horizontal web stiffeners, the full shear area approach should be used (Figure 27a). The values of My presented in the Table 11 have been computed using both approaches, but only the full shear area approach is close to the maximum moment obtained from the tests (in the case of anti-symmetrical loading).

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Table 11 Comparison of test to the analytical results given by Annex J of EC3 for steel joints Total Energyφmax

+ φmax- Mmax Mmin Sj,ini

+ Sj,ini- φy

+ φy- My

+ My-

Specimen kNm rad mrad kNm x103kNm/rad mrad KNm Symmetrically loaded joints

EC3-full As --- --- --- 55.64 2.97 165.40 EC3–red. As --- --- --- 55.64 2.97 165.40 BX-SS-M 9.01 43.20 263.34 48.03 3.26 180.79 BX-SS-C1 41.5 28.0 21.0 271.6259.1 55.91 59.60 3.26 2.60 197.2 188.0BX-SS-C2 23.4 17.4 18.1 261.8259.8 71.24 63.5 2.66 2.39 194.8 206.8

Anti-symmetrically loaded joints EC3-full As --- --- --- 32.99 4.76 156.91 EC3–red. As --- --- --- 25.19 4.24 106.77 BX-SU-M 24.0 105.5 258.36 51.50 2.28 137.66 BX-SU-C1 135.8 72.5 55.3 269.4240.6 35.07 29.08 3.77 4.44 153.1 161.2BX-SU-C2 88.37 39.2 46.8 240.1236.6 27.82 40.53 5.54 3.37 179.8 161.2full As – full shear area approach

red As –reduced shear area approach, according to Annex J of EC3

shear areaFULL SHEAR

AREA SUPPLEMENTARY WEB PLATE

shear area

(a) (b) Figure 27: Full shear approach (a) and EC3 Annex J approach (b)

The experimental curves of the monotonic tests, as well as the envelopes of the cyclic tests, are presented in Figure 28, compared to the analytical results given by the Annex J of Eurocode 3. In the case of symmetrically loaded joints, it can be observed that the values of the experimental stiffness are close to those computed by Annex J, but the values of the computed yielding bending moment is generally 10-20% smaller than the experimental ones. For the anti-symmetrically loaded joints, only the full-shear area approach gives close agreement to the analytical results given by Annex J. The monotonic test is different from the cyclic tests in stiffness, maximum resistance and maximum rotation, as can be seen in Table 11. As usual, the maximum rotation in monotonic tests is 1.5-2 times greater than the maximum rotation attained under cyclic loading. The conventional values of yielding moment and of yielding rotation for the case of load reversal are closer to the computed values by Annex J.

BX-SS

-350

-250

-150

-50

50

150

250

350

-0.12 -0.09 -0.06 -0.03 0.00 0.03 0.06 0.09 0.12TOTAL JOINT ROTATION [rad]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

] BX-SU

-350

-250

-150

-50

50

150

250

350

-0.12 -0.09 -0.06 -0.03 0.00 0.03 0.06 0.09 0.12TOTAL JOINT ROTATION [rad]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

EXP MON. EXP CYCLIC THEORETICAL (EC3)

ECCS-TC10-03-528

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Figure 28: Comparison of experimental results for steel joints to the EC3 Annex J predictions. Composite joints Table 12 shows the test results, compared to the numerical ones given by Annex J. For the case of composite joints, the full shear area of the panel zone has been considered, too. Figure 29 shows graphically the same comparison (symmetrical loading - a, anti-symmetrical loading - b). In the case of cyclic loading are shown only the envelope curves. The shear connection was computed according to EC4, part 1.6 in order to have a full-shear connection between the concrete slab and the steel beam. The studs were distributed according to the shear force pattern under gravitational loads. Table 12 Comparison of test to the analytical results (Annex J of EC4) for composite joints

Total Energyφmax+ φmax

- Mmax Mmin Sj,ini+ Sj,ini

- φy+ φy

- My+ My

-

Specimen kNm rad mrad kNm x103KNm/rad mrad KNm Symmetrically loaded joints

EC4 Mom - --- --- --- 57.11 2.34 133.80 EC3* Mom + --- --- --- 99.68 2.30 230BX-CS-M1 16.60 86.7 244.25 98.47 1.27 155.51 BX-CS-M2 8.4 32.7 316.09 105.27 2.52 231.32 BX-CS-C1 32.0 44.8 14.1 305.5196.9 102.5 75.05 1.60 1.73 195.2 149.9

Anti-symmetrically loaded joints EC4-comb.** --- --- --- 41.78 3.84 160.55 BX-CU-M 15.2 90.1 198.24 47.19 2.34 126.23 BX-CU-C1 104.5 60.5 53.8 187.1193.3 36.87 37.92 3.49 3.38 142.67 137.3BX-CU-C2 32.2 32.7 30.9 191.3210.5 -- -- -- -- -- --

EC3* - computed according to Annex J of EC3, by translation of the centre of compression EC4-comb.** - mean value between the positive and negative values

The analytical computations according to Annex J of EC4 in the case of symmetrical positive bending leads to safer values in terms of resistance (15% approx.) and smaller stiffness. Due to slab degradation in cyclic tests, the resistance and stiffness values are smaller. The Annex J of EC4 does not give the possibility of computing composite joints subjected to positive moments. For this case, the values of stiffness and moment resistance presented in Table 12 are computed according to Annex J of EC3 by a translation of the centre of compression from the upper beam flange to the middle of the concrete slab (considered without corrugated sheet). These assumptions lead to comparable values to the tests in terms of resistance and stiffness. For the case of cyclic loading, there can be observed a decrease in both resisting moment and stiffness due to rapid slab degradation. Analytical values of moment resistance and stiffness for anti-symmetrical loading have been obtained by the mean values for the two connections subjected to anti-symmetrical loading, taking into account the full shear area approach. This prediction remains only an attempt of computing the composite joints under anti-symmetrical loading.

ECCS-TC10-03-528

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BX-CS

-350

-250

-150

-50

50

150

250

350

-0.12 -0.09 -0.06 -0.03 0.00 0.03 0.06 0.09 0.12TOTAL JOINT ROTATION [rad]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

] BX-CU

-350

-250

-150

-50

50

150

250

350

-0.12 -0.09 -0.06 -0.03 0.00 0.03 0.06 0.09 0.12TOTAL JOINT ROTATION [rad]

MO

MEN

TA

TTH

EC

OLU

MN

FAC

E[k

Nm

]

EXP MON. EXP CYCLIC THEORETICAL (EC4) Figure 29: Comparison of experimental results for composite joints to the EC4 Annex J predictions 6.4. General conclusions and remarks The use of X-shaped columns makes possible a convenient design for three- and four- way connections for space moment resisting frames. Also, it brings important advantages to the joint behaviour under anti-symmetrical loading over usual I or H shaped columns. Column flanges parallel to the considered web lead to a natural stiffening of the column panel zone. This increase in the panel zone shear area reduces significantly the drop in moment capacity for anti-symmetrically loaded joints with respect to symmetrical ones, but reduces to some extent the initial stiffness. Anyway, the stiffened panel zone participates to the plastic mechanism, assuring a significantly increased ductility of anti-symmetrically loaded joints. Cyclic loading introduces differences between the type of failure of both bare steel and composite joints. While for monotonic tests the failure was mainly by bolt failure and column flange / end plate deformations, in the case of cyclic tests it was by brittle failure of the fillet welds. Therefore, particular care is needed in design and manufacture of the welds in zones with load reversals. Full-penetration welds could be more reliable. However, weld quality is of paramount importance. The cyclic loading reduces the ductility of bare steel and composite joints. Roughly, the maximum rotation of cyclically loaded joints is 50% of the monotonically loaded ones for bare steel joints and anti-symmetrically loaded composite joints, and 33% for symmetrically loaded composite joints. However, in the case of anti-symmetrically loaded joints (which is the case under seismic loading), the plastic rotation capacity is greater then the generally accepted requirement of 0.03 rad. for special MRFs (AISC 1997). This is valid both bare steel and composite joints considered in this study. The maximum bending moment maxM is affected by the cyclic loading in the case of composite connections only. A 10% reduction of the maximum bending moment attained under monotonic loading could be considered as a safe estimation of the maximum bending moment under seismic loading. Composite action of the concrete slab on the steel beam has a positive effect on the ductile behaviour of the symmetrically loaded joints under negative moments. Under positive moments, the centre of compression is shifted into the concrete slab, leading to a higher lever arm for the extreme bolt rows, and higher stiffness and moment capacity of the joint. Consequently, in the case of composite joints under positive moments, the lower steel part of the joint should be proportioned accordingly, in order to resist the increased demand due to composite action. In the case of composite joints with partially extended end plate (at the bottom part only), the steel reinforcement in the slab is not able to compensate for the missing bolt row in the extended end plate, even if a relatively high reinforcement ratio is used (1.8% for the joint zone). A pinching behaviour could be observed for the anti-symmetrically loaded specimens, after cracking of the concrete slab. A higher strength concrete is believed to postpone this phenomena, and to result in more stable behaviour. Due to rapid degradation of the concrete slab in the case of anti-

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symmetrically loaded composite joints and pinching behaviour, the dissipated energy per cycle is relatively low. A total shear connection between the concrete slab and the beam as defined by EC 4 (ENV 1994-1-1 1992) leads to an adequate performance of the connectors in the joint zone. There have been no recorded failures of the connectors in tests performed. Analytical model of Annex J - EC3 for steel joints and EC4 for composite joints respectively - provides a reliable prediction for behaviour of I beam to X shaped columns with extended end plate connections, but an appropriate modeling of the panel zone should be used. When transverse stiffeners are used, the effective shear area of the X-shaped column panel zone should be considered as the sum of the shear areas of the column web and the two flanges parallel to the considered web (Dubina et al, 2000b). The experimental results of composite joints and the EC 4 Annex J predictions show a good agreement, the code offering a good basis for design of composite joints under negative bending. For design of composite joints under positive bending, Annex J provisions could be used, but taking the centre of compression shifted into the concrete slab. In case of anti-symmetrical loading an adequate modeling should be found. New modern seismic codes (AISC 1997), based on experimental proofs of structural beam-to-column joints, classify the joints by different levels of ductility and resistance but they do not offer the verification tool for structural analysis. The joint modeling that is permitted by element 14 of DRAIN 2DX, accompanied by similar models for structural members can offer a solution to this issue. These models give results close to reality, for an accurate structural analysis, especially in the post-elastic domain where the quantity and location of energy dissipation becomes important. In the case of a design based on performance criteria (Bertero, 1997), such modeling could be considered as the controlling tool for the required criteria.

7. General Concluding Remarks The values of moment resistance Mu and initial rotational stiffness Sj,ini of the joints subjected to cyclic loading are remaining practically the same as the values obtained for the same joints but subjected to monotonic loading. Concerning the rotation capacity of the joints under cyclic loading, this very important characteristic could be evaluated with a good approximation as half of the rotation capacity determined on the joints under monotonic loading. It is clear for all specialists of the field that the rotation capacity of beam-to column joints is one of the main characteristics which are influencing the seismic behaviour of the steel MR frames. That is the reason why in the last period, this characteristic has been introduced with some arbitrary values of reference for different types of frames. Unfortunately, formulae for evaluating this characteristic are very few. It is evident that the researches have to be continued in both directions experimental test research and numerical modelling, to establish new definitions for the evaluation of the rotation capacity of beam-to-column joints.

References 1. Eurocode 3. 1992. Design of steel structures. Part 1-1: General rules and rules for buildings.

European prestandard. 2. Grecea D. 1999. Caractérisation du comportement sismique des ossatures métalliques -

Utilisation d’assemblages à résistance partielle. Thèse de Doctorat. INSA de Rennes, France. 3. Aribert, J.M. & Grecea, D. 1997. A new method to evaluate the q-factor from elastic-plastic

dynamic analysis and its application to steel frames. Proceedings of STESSA’97, Kyoto, 3-8 August 1997.

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4. Aribert, J.M., Dubina, D., Grecea, D. & Dinu, F. 1997. Parametrical study on a new method for q-factor evaluation. Proceedings of STESSA’97, Kyoto, 3-8 August 1997.

5. Aribert, J.M. & Grecea, D. 1998. Experimental behaviour of partial-resistant beam-to-column joints and their influence on the q-factor of steel frames. The 11th European Conference of Earthquake Engineering, Paris, 6-11 September 1998.

6. Aribert, J.M. & Grecea, D. 2000. Numerical investigation of the q-factor for steel frames with semi-rigid and partial-strength joints. Proceedings of STESSA’2000, Montreal, 21-24 August 2000.

7. Revised Annex J of Eurocode 3, Joints in Building Frames, Edited approved draft, CEN Document CEN/TC 250/SC 3 – N 671 E, January 1997.

8. Dubina, D., Grecea, D. & Dinu, F. 1997. ESDEP WG14: Structural systems: Buildings. Lecture 14.13: Design of Multi-Storey Frames with Partial Srength and Semi-Rigid Connections, WIVISS Wider Vocational Initiative in Structural Steelwork, 1997.

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11. AFNOR-Règles PS 92 appliquables aux bâtiments, NFP 06.013, Décembre 1995. 12. Uniform Building Code, Volume 2, Structural Engineering Design Provisions. International

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Ministry of Public Works and Territory Planning, Romania, 1992. 16. Kuhlmann, U. & Kuhnemund, F. 2000. Rotation capacity of steel joints, NATO Advanced

Research Workshop “The Paramount Role of Joints into the Reliable Response of Structures, From the Rigid and Pinned Joints to the Notion of Semi-rigidity”, Ouranoupolis, Greece, 21-23 May 2000.

17. Ciutina A., Stratan A. 1999. Cyclic tests on beam to column connections. Second international conference of PhD students, Miskolc, Hungary

18. Eurocode 3 Part 1.1 1992. General rules and rules for buildings. CEN, Brussels, Belgium 19. ECCS, 1986. Recommended Testing Procedures for Assessing the Behaviour of Structural

Elements under Cyclic Loads, European Convention for Constructional Steelworks, Technical Committee 1, TWG 1.3 – Seismic Design, No.45

20. SAC Joint Venture 1995. Connection Test Summaries. Report No. SAC-96-02, Sacramento, California, USA

21. SR EN 10002-1 1990. Metallic materials tensile testing. CEN, Brussels, Belgium 22. Suita K., Nakashima M., Morisako K. 1998. Tests of welded beam-column subassemblies.

Journal of structural engineering. November, 1236-1252 23. AISC 97 (1997). Seismic Provisions for Structural Steel Buildings. American Institute of Steel

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25. Ciutina, A, Stratan A., Dubina D. 2001 “Răspunsul seismic al cadrelor metalice multietajate în funcţie de modelele M-Φ utilizate pentru îmbinări şi elemente structurale.” Proceedings of the Second National Conference on Earthquake Engineering - UAICR, Bucureşti, 8-9 November, 2001, Vol 2, 3.113-3.123.

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26. Dubina, D., Grecea, D., Ciutina, A., Stratan, A. (2000). "Influence of Connection Typology and Loading Asymmetry", Chapter 3.2 in: Moment Resistant Connections of Steel Building Frames in Seismic Areas -, (Mazzolani F.M. ed.) E&FN SPON (London).

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28. Dubina, D., Ciutina, A., Stratan, A. (2001). "Cyclic Tests of Double-Sided Beam-to-Column Joints", Journal of Structural Engineering, Vol.127, No.2, Feb.2001, pp.129-136

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