vancouver convention centre expansion project...

VANCOUVER CONVENTION CENTRE EXPANSION PROJECT – GEOTECHNICAL ASPECTS OF DESIGN AND CONSTRUCTION M. Yogendrakumar, Golder Associates Ltd., Burnaby, B.C., Canada Upul Atukorala, Golder Associates Ltd., Burnaby, B.C., Canada Neil Wedge, Golder Associates Ltd., Burnaby, B.C., Canada Humberto Puebla, Golder Associates Ltd., Burnaby, B.C., Canada Viji Fernando, Golder Associates Ltd., Burnaby, B.C., Canada Norm McCammon, Golder Associates Ltd., Burnaby, B.C., Canada ABSTRACT Construction of the Vancouver Convention Centre Expansion is currently underway. The 106,000 m2 expansion to the west of Canada Place will include construction of a new convention centre, viaduct, a harbour concourse linking the existing convention centre, and a unique six acre environmentally friendly “living roof”. This paper describes the history of the site development and its impact on the selection of the foundation system for this project. Results of the state-of-the-art dynamic soil-structure (FLAC) analyses and the test pile and test densification programs completed to support the foundation design are discussed. The results and lessons learned from the construction phase to date are also discussed. RÉSUMÉ Les travaux de construction relatifs à l’expansion du Centre des congrès de Vancouver sont présentement en cours. Le projet d’expansion, qui s’étend sur 106 000 m2 à l’ouest de la Place du Canada, inclura la construction d’un nouveau Centre des congrès, d’un viaduc et d’un quai relié au centre déjà existant ainsi que d’un grand « toit vert » de 6 acres unique en son genre. Le présent document décrit l’historique du développement du site et son impact sur le choix du système de fondations retenu. Des résultats sont aussi présentés au niveau de l’analyse dynamique de l’interaction des structures et des sols à la fine pointe des connaissances. L’analyse des empilements et des essais de densification réalisés pour soutenir le modèle de conception choisi est aussi présentée. Les résultats obtenus et les leçons apprises à ce jour au niveau de la construction sont aussi discutés. 1. INTRODUCTION The C$615 million Vancouver Convention Centre Expansion Project (VCCEP) is expected to be completed in the fall of 2008. In 2010, the facility will welcome some 10,000 media members who are expected to attend the 2010 Olympics and Paralympic Winter Games. Artist rendering of this facility is shown in Figure 1.

Figure 1. Artist Rendering of VCCEP Facilities This expansion and upgrade of the 18 years old existing Vancouver Convention and Exhibition Centre (VCEC) will allow Vancouver to attract conventions of more than 5,000 delegates. The 106,000 m2 expansion to the west of Canada Place will include construction of a new convention centre, viaduct, a harbour

concourse linking the existing convention centre, and a unique six acre environmentally friendly “living roof”. The project site is located in Coal Harbour between Canada Place to east and the Harbour Green Park to the west. The site is shown in Figure 2.

Figure 2. Aerial Photograph of VCCEP Site

SITE

Sea to Sky Geotechnique 2006

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2. SITE DEVELOPMENT HISTORY The modern development of the Coal Harbour area started with the establishment of the western terminus of the Canadian Pacific Railway (CPR) in this area in the 1880’s followed by port development and subsequent industrial development along the foreshore. An aerial photograph taken in 1946, presented in Figure 3, shows a shipping pier (then known as Pier A) projecting into Coal Harbour at the now proposed VCCEP site. It was located just west of a similar but larger pier (Pier B-C, now called Canada Place). The VCCEP site pier (Pier A) was still operating in the air photos taken in 1954 and 1963 (not

shown) along with another, newer pier called Pier A3 west of Pier A. However, a 1968 photograph shows that land had started to be reclaimed by placing fill on the west side of Pier A following demolition of the western apron of the pier. The remnant of Pier A (eastern apron) was subsequently referred to as Pier A2. The reclamation fill had been extended to approximately its current plan shape by 1976, some 30 years ago. By 1976, a new pier called Pier A1 located between Pier A3 to the west and Pier A2 to the east was operating. Remnants of an old stone seawall are still visible below the West Quay wharf which once connected the western apron of the Pier B-C (now called Canada Place) with the eastern apron of Pier A. The western extent of the remnants of this old stone seawall was not known.

Figure 3. Aerial Photographs of the Site An oblique aerial photograph taken on May 2003, at or about the time of project inception, shows the use of the site as a commercial marina and floatplane terminal (see Figure 2). Modification and improvement to the Coal Harbour shoreline to the west of the VCCEP site have been carried out by Marathon Developments Inc. in several phases. The latest reclamation and shoreline development to the

immediate west and in the northwest portion of the proposed Vancouver Convention Centre Expansion site were completed in 2000. 3. GENERAL SUBSURFACE CONDITIONS A number of geotechnical boreholes have been drilled and sampled in both the onshore and offshore areas of

1954 1946

1968 1999

Pier B-Pier A

Pier A2

Slope

Pier A3

Pier A1 Pier A3


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the proposed VCCEP site for various development projects (see Figure 4). The inferred stratigraphy along the approximately north-south axis of the foundation footprint (Section A-A) is shown on Figure

5. The four main stratigraphic units encountered at this cross-section, in order of increasing depth, are as follows:

Figure 5. Inferred Stratigraphic Section A-A’ 3.1 Man-made Fills As noted under site history above, the Vancouver Convention Centre Expansion site comprises largely of land that has been reclaimed over the years. The thickness of the man-made fill materials is generally least near the southern edge of the site (about 9 m) and increases northward, as the water deepens, to the current shoreline where it is about 21 m thick. Based on information obtained from CP Rail Archives, the fills used in the construction of Pier A were retained by rock toe berms on the east and west sides that were oriented perpendicular to the then shoreline. The source of the Pier A core fills is unknown, however, these fills are primarily sand and gravel with trace to some silt and are believed to be specially sourced fills rather than random fills. Based on information available from the CP Rail archives, timber piles were driven through the fills to support the eastern and western aprons of Pier A and the railway track trestle ran in the middle portion of Pier A. The fills of the 1968 reclamation were reported to be random fills and, based on borehole information, these fills are heterogeneous in nature and consist of silts, sands and gravels with varying amounts of brick fragments, cobbles, boulders, woodwaste, concrete rubble and other materials. 3.2 Marine Sediments Beneath the fills, where they are present, and at the seabed in the areas offshore of the fills, are the original marine seabed deposits covering this area. These range in thickness from about 3 m in the southern part of the site to about 17 m offshore and consist of compressible silts and clayey silts (with some organics) to sand and silty sand (with shells and gravel). The coarser-grained soils are inferred to be near surface beach deposits. The finer marine sediments are firm to stiff in consistency below the onshore area, and grade into a very soft to soft

consistency some distance offshore of the crest of the slope. 3.3 Glaciomarine Deposits and Glacial Drift Underlying the marine sediments, except in the southern portion of the site, is a stratum of silt, sand, gravel and cobbles inferred to be glaciomarine and glacial drift deposits. The relative density of this deposit varies from compact to very dense in the upper 10 m thickness and dense to very dense at depth. The relative density is variable at any given elevation or depth. The thickness of this deposit, where present, ranges from about 6 to 27 m. 3.4 Sedimentary Bedrock Bedrock consisting of interbedded sandstone, siltstone and mudstone (i.e. sedimentary rocks) was encountered across the site at depths ranging from about 12 m below ground surface at the south side of the site increasing to about 43 m below seabed at the north side. The rock is in various states of weathering from completely weathered to slightly weathered across the site, and is classified generally as very weak to weak. 4. FOUNDATION SCHEME At the concept design stage, two major foundation schemes for support of the VCCEP facilities were evaluated. One of the schemes consisted of drilled shafts varying in diameter from 1.2 to 3 m which would be socketted into the underlying bedrock. The other foundation system consisted of 914 mm diameter vertical steel pipe piles driven into the dense glacial drift and/or sedimentary bedrock. In November 2003, a detailed study by Westmar Consultants Inc., the marine structural consultant for


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the project, concluded that a piled structure would be more economical than the drilled shaft option. An open ended piled foundation, even with provision for obstructions during driving and embedment into glacial drift to derive the actual capacity, is considered to be the lowest risk option. This study, which was peer reviewed by two external consultants, recommended a test pile and test densification program to obtain information for the designers and contractors bidding on this work. 5. TEST PILE & TEST DENSIFICATION

PROGRAMS The primary purposes of the test pile, test densification and test pit program were to:

• verify the ability to densify/strengthen the site fills and underlying marine deposits to the target densification requirements;

• determine/confirm the offset (window) between proposed test pile and stone column treatment points required to avoid a significant increase in the level of effort required for the pile installation;

• determine the pile embedment requirements within the glacial drift or sedimentary bedrock bearing strata to achieve the design loads;

• determine/confirm the energy requirements of the hammer to install the piles to achieve the design loads;

• determine the feasibility and effectiveness of installation of piles through existing densified soils;

• obtain load-settlement characteristics of the installed test piles;

• verify the existence of the remnant foundation piling and other elements associated with the original Pier A development; and

• verify the extent of the remnants of the old stone seawall and general composition of the old shoreline.

5.1 Test Densification Two test areas, one on the west side (Area 4) and one on the east side (Area 5) of the convention centre site, were selected for ground improvement treatment by installation of stone columns using vibro-replacement techniques. The locations of the two test treatment areas are shown in Figure 4. An electric poker type V-23 vibroflot was used for the ground improvement work. A triangular pattern of stone columns at a centre to centre spacing of 2.75 m was selected as a means of achieving the desired level of ground improvement. Mud rotary drilling and Standard Penetration Testing were carried out to assess the degree of ground

improvement achieved. Results of the tests carried out to verify the degree of densification achieved following stone column installation are shown in Figure 6 for the east side. The test densification program indicated that the site fills and the underlying marine deposits could be densified/strengthen to the target levels.

Figure 4. Site Plan 5.2 Test Pile Program The test pile program was carried out by FRPD. A total of five test piles (P1 to P5) were driven at the locations shown in Figure 4. The pile driving and testing was carried out during the period between February 25, 2004 and March 31, 2004. The test piles were driven using an APE D80-23 diesel hammer with a maximum rated energy of about 267 kJ/blow. Test pile P1 was driven in a previously densified area where stone columns were installed as part of the Phase II shoreline development by Marathon Development Inc. The purpose of this test pile was to assess driveability through densified ground containing 50 mm minus stone backfill. Test piles P4 and P5 were driven in the test treatment Areas 4 and 5 following completion of the test densification program. The purpose of these piles was to assess driveability through the rock toe berms placed during construction of Pier A. Test piles P2 and P3 were driven near the current crest of the shoreline. All test piles were driven open-ended, however, the piles were fitted with an inside cutting edge/shoe manufactured by Associated Pile & Fitting Corporation (Fitting Number 0-14001). The test piles were vibrated to a suitable depth to free stand using an APE model 300 vibro hammer prior to driving with the APE D80 diesel hammer. Pile driving analyzer (PDA) testing was carried out during the pile driving at the end of initial driving


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(EOID) and at re-strike. PDA testing at re-strike was not carried out for test pile P3 due to schedule constraints and lack of pile stick-up following driving to the final tip elevation.

Figure 6. Test Densification Results for East Area All test piles were subjected to Statnamic testing to obtain their load-displacement characteristics. Figure 7 shows the completed set up for the Statnamic testing at Test pile P5. A summary of the PDA and Statnamic Testing is provided in Table 1 below: Table 1. Summary of Load Test Data

The test pile program indicated that the piles can be driven open ended into glacial drift and sedimentary bedrock to achieve an allowable compression capacity of 4500 kN. The required pile embedment into glacial

drift and sedimentary bedrock to achieve the above-noted geotechnical capacity was about 20 m and 3 m respectively. The minimum rated energy of the pile hammer required is determined to be about 267 kN/blow.

Figure 7. Statnamic Load Testing Set-up The load-displacement curve obtained from Statnamic Testing for Pile no. P3 is shown in Figure 8.

Figure 8. Load Displacement Response of Test

Pile P3

6. DYNAMIC SOIL STRUCURE INTERACTION ANALYSES

One of the key challenges faced by the designers was to assess the extent of soil liquefaction and the

** Mobilized Capacity from PDA Tests (kN)

Statnamic Load Test Pile No.

At EOID

At Re-strike

Maximum Test Load

(kN)

Maximum Vertical

Deflection (mm)

Residual Deflection

(mm)

P1 6,000 n/a 8,334 35 11 P2 8,200 8,200 9,232 34 2 P3 6,600 * n/a 11,211 42 10 P4 10,200 13,000 8,284 21 5 P5 9,500 9,500 10,102 29 5


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resulting impact on the pile foundations under the design (1:475-year) seismic loading conditions. Initial assessments were carried out using simplified methods of analysis based on average penetration resistance measurements. The results of these analyses indicated that the loose to compact heterogeneous man-made fill materials and portions of the marine sediments have a high risk of liquefaction under design ground motions corresponding to a peak horizontal ground acceleration of 0.2 g. The resulting peak transient and permanent lateral ground displacements in the fill materials and foundation structure movements at main deck level needed to be evaluated as input to the substructure design. Analyses were initially undertaken using simplified and uncoupled methods where the response of the individual piles was analyzed using the computer code LPILE and imposing the lateral free-field displacements estimated using empirical models such as Youd et al (1999). In these analyses, the soil reaction developed at the soil-pile interface was modelled using “p-y” curves derived for both liquefied and non-liquefied/competent soils. These analyses indicated foundation movements to be excessive at deck level. In light of the increasing thickness of fill materials present at the site towards water, the likely differences in the time at which liquefaction will be triggered in the different areas of the soil profile and extent of soil liquefaction, and the anticipated complex interaction response between the soil and foundation piles, it was considered necessary to develop coupled soil-structure interaction models to assess the seismic loading-induced deformations of the substructure to permit a displacement-based design. It was also recognized that it was important to incorporate the superstructure inertial loads at least in an approximate manner in these analyses. Coupled soil-structure interaction models were developed to assess the effects of design ground shaking using the finite difference computer code FLAC2D. The foundation piles were incorporated using the “pile element” option and the soil-pile interface response was modelled using shear and normal soil springs attached to the grid nodes. A simplified structural frame model that captures the non-linear stiffness and mass of the superstructure was added to the substructure to take into consideration the effects of the inertia loads originating from the superstructure on the substructure.

The user-defined stress strain module UBCTOT developed by Dr. Peter Byrne and his colleagues at the Department of Civil Engineering, University of British Columbia was used to simulate the pre-liquefaction and post-liquefaction stress-strain-strength response of potentially liquefiable soils. UBCTOT permits sequential modelling of soil liquefaction in real time of each soil zone as and when a prescribed set of conditions are met; i.e. number of equivalent cycles for liquefaction for the design magnitude event. Upon triggering of liquefaction, the shear strength and shear stiffness of the zone comprising the liquefiable soils are assigned new parameters indicative of liquefied soils. Details of this user-defined stress-strain module are given elsewhere (Beaty and Byrne, 1999). Two separate FLAC models were developed because of the differences in the structural model between the east and west halves of the proposed structure and the incorporation of an expansion joint separating the structures, differences in soil conditions, and the presence of previously improved ground in the eastern half of the site. The finite difference models developed for the two areas are shown in Figures 9 and 10. Models only along the north-south orientations were considered because this direction constitutes the weakest direction of the structure with respect to seismic loading. Typical soil parameters used to represent the different soil zones are summarized in Tables 2 and 3 for the east and west models, respectively. Both free-field (i.e. without structure) and coupled soil-structure interaction models were developed. Also, analyses were undertaken both with and without ground improvement measures to assess the relative impact of ground improvement on foundation movements and forces and moments developed in the piled foundations. Based on structural considerations, it was considered necessary to densify the soil in the upper several pile diameters in order to provide the necessary soil confinement to reduce the development of large bending moments at deck level. A 30 m wide strengthened perimeter seismic dike extending from surface all the way to the top of glacial drift was considered to improve the lateral stability of the site and to minimize the seismic loading-induced lateral movements of the overall site in general, and the perimeter areas in particular.


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Figure 9. Finite difference east model

Vancouver Convention Centre Expansion Project

Bedrock

Glacial Drift

Marine Silt

Unimproved Heterogeneous Fill

Densified Heterogeneous Fill

Pier A Fill

Marine Silty Sand

(north)

Marine Silty Sand

(south)

Fill and Silty Sand

Under Viaduct

R Q P N M L K J H G F E D C B A.3

FLAC (Version 4.00)

LEGEND

19-Apr-04 15:05 step 25522 -1.664E+01 <x< 2.762E+02 -1.803E+02 <y< 1.126E+02

(N1)60-cs 0.000E+00 1.000E+00 1.000E+01 1.900E+01 2.000E+01 2.500E+01 2.700E+01 2.900E+01 5.900E+01

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0.250 0.750 1.250 1.750 2.250(*10^2)

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Figure 10. Finite difference west model Table 2. Summary of soil parameters – East Model

Soil Unit γsat (kN/m3)

(N1)60 (blows/0.3m)

φ (deg)

Su (kPa)

Vs (m/s)

Gmax (MPa)

ν

Heterogeneous Fill (Viaduct)

18 10 34 - - 94 (σ’m/Pa)1/2 0.45/0.33*

Pier A Fill 20 27 36 - - 130 (σ’m/Pa)1/2 0.45 Densified Fill 20 20 36 - - 118 (σ’m/Pa)1/2 0.45 Southern Silty Sand

18 17 35 - - 112 (σ’m/Pa)1/2 0.45

Northern Silty Sand

18 17 35 - - 112 (σ’m/Pa)1/2 0.45

Marine Silt 17.4 - - 0.30 σ’v ≥ 10 kPa

- 1,300 Su 0.45

Reinforced Marine Silt

17.4 - - 0.37 σ’v

≥ 10 kPa - 1,300 Su 0.45

Glacial Drift 20 - 38 - - 169 (σ’m/Pa)1/2 0.33 R0 Bedrock 20 - 45 - 450 (γ/g)Vs

2 0.125 R1 Bedrock 20 - 45 - 800 (γ/g)Vs

2 0.125 Bedrock 20 - 45 - 1200 (γ/g)Vs

2 0.125 Notes: * Fill material above water Table. γsat is the unit weight; (N1)60, the normalized standard penetration resistance; φ, the friction angle; Su, the undrained shear strength; Vs, the shear wave velocity; Gmax, the small strain shear modulus; ν, the Poisson’s Ratio; σ’m, the mean effective stress; σ’v, the vertical effective stress and g, the gravity acceleration.

Vancouver Convention Centre Expansion Project

Bedroc

Glacial Drift

Marine Silt

Phase 2Dense Fill

Densified Heterogeneous Fill

Marine Silty SandWeathered

Bedrock

Fill and Silty Sand

Under Viaduct

R Q P N M L K J H G F E D C B A.3

FLAC (Version 4.00)

LEGEND

10-May-04 18:48 step 26549 -1.189E+01 <x< 2.260E+02 -1.409E+02 <y< 9.694E+01

Material_Ind 1.000E+01 1.700E+01 2.000E+01 3.000E+01 3.700E+01 4.000E+01 4.500E+01 6.000E+01 6.500E+01 7.000E+01

Grid plot

0 5E 1

Pile plotBeam plotCable plot -1.200

-0.800

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Table 3. Summary of soil parameters – West Model

Soil Unit γsat (kN/m3)

(N1)60 (blows/0.3m)

φ (deg)

Su (kPa)

Vs (m/s)

Gmax (MPa)

ν

Heterogeneous Fill (Viaduct)

18 10 34 - - 94 (σ’m/Pa)1/2 0.45/0.33*

Phase 2 Fill 20 40 37 - - 148 (σ’m/Pa)1/2 0.45 Densified Fill 20 20 36 - - 118 (σ’m/Pa)1/2 0.45 Silty Sand 18 17 35 - - 112 (σ’m/Pa)1/2 0.45 Marine Silt 17.4 - - 0.30 σ’v

≥ 10 kPa - 1,300 Su 0.45

Reinforced Marine Silt

17.4 - - 0.37σ’v ≥ 10 kPa

- 1,300 Su 0.45

Glacial Drift 20 - 38 - - 169 (σ’m/Pa)1/2 0.33 R0 Bedrock 20 - 45 - 450 (γ/g)Vs

2 0.125 R1 Bedrock 20 - 45 - 800 (γ/g)Vs

2 0.125 Bedrock 20 - 45 - 1,200 (γ/g)Vs

2 0.125 Notes: * Fill material above water Table. γsat is the unit weight; (N1)60, the normalized standard penetration resistance; φ, the friction angle; Su, the undrained shear strength; Vs, the shear wave velocity; Gmax, the small strain shear modulus; ν, the Poisson’s Ratio; σ’m, the mean effective stress; σ’v, the vertical effective stress and g, the gravity acceleration. Typical results showing the computed extent of soil liquefaction and the structural movements are shown in Figure 11. The analyses indicated that the substructure would likely experience transient and permanent movements in the order of 80 to 115 mm and 10 to 15 mm, respectively. Typical displacement time-histories computed at deck level (of the east structure) are shown in Figure 12. The computed maximum transient bending moments and axial compression forces in the pile elements were in the order of 2,500 kN-m/pile and 9,400 kN/pile. These deformations, bending moments, and shear forces were considered acceptable from a structural point of view.

7. CONSTRUCTION UPDATE 7.1 Ground Improvement The ground improvement work began in December 2004. Because of the difficulties encountered with waste water sediment control facilities during the test densification program, the dry bottom feed technique was generally employed with provision for water jetting to overcome obstructions, etc. S-type vibrators were used in the stone column installation. The pier A fills responded very well to the ground improvement work. As expected, difficulties primarily related to obstructions were encountered within the heterogeneous site fills. These challenges were overcome by various means including pre-augering, relocation and changes to the vibrator units.


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Figure 11. Typical pattern of liquefaction extent and structural movements A total of about 39 km and 23 km of stone columns were installed in the onshore and offshore areas of the site, respectively. The total number of stone columns installed amounted to about 3525 and 2100 in the onshore and offshore areas, respectively. More details regarding ground improvements are given in a companion paper. 7.2 Pile Installation The pile installation for the convention centre structure consisted of 914 x 19 mm steel pipe piles driven open ended. The piles were fitted with an inside cutting shoe. The piles were driven using a Birmingham B6505 diesel hammer or a B6505HD hammer with

rated energies of 274 and 344 KJ/blow, respectively. A total of 889 piles were installed representing a total installed pile length of approximately 30 km. As expected, several piles encountered refusal at shallow depths within the man-made fills. These are primarily in the 1968 reclamation areas and near the current shoreline. In some cases, piles and pile tips were severely damaged requiring the piles to be pulled out and obstructions removed by excavation. A pile that was severely damaged within the heterogeneous site fills is shown in Figure 13.

Vancouver Convention Centre Expansion

Liquefied

FLAC (Version 4.00)

LEGEND

28-Jul-04 4:25 step 2209217 -1.196E+01 <x< 2.268E+02 -1.414E+02 <y< 9.738E+01

Liq_Ind 0.000E+00 1.000E+00 2.000E+00 8.000E+00 9.000E+00

Grid plot

0 5E 1

Pile plot Exaggerated Disp.Magnification = 2.500E+01Max Disp = 1.079E-01Beam plot Exaggerated Disp.Magnification = 2.500E+01Max Disp = 3.607E-01Cable plot

-1.200

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Figure 12. Displacement time-histories at deck level Selected piles were subjected to dynamic testing using Pile Driving Analyzer (PDA) to confirm axial compression capacities. A 40 kip (20 ton) drop hammer was used at re-strike condition to prove axial compression capacity since the initial phase of PDA testing suggested that the Birmingham B6505 hammer was not able to fully mobilize the piles. With the 40 kip drop hammer, the PDA testing was able to prove ultimate compression capacities as high as 13,000 kN For the southern piles, which were founded in the sedimentary bedrock, the embedment into the bedrock was found to be generally in the order to 2 to 3 m. The northern piles were generally driven about 20 m into the glacial drift in order to achieve an ultimate compression capacity of 10,000 kN. Acknowledgements The authors would like to thank VCCEP Ltd. for their permission to publish this paper. Several individuals and consulting companies were involved in this project and their input to the formulation of the geotechnical design and construction concepts are acknowledged. In particular, we would like to acknowledge the

contributions of Westmar Consultants Inc. We would also like to acknowledge the efforts of many speciality contractors. Their contributions were vital for the successful completion of this interesting and challenging project.

Figure 13. Damaged Pile being extracted

Vancouver Convention Center Expansion Project

Bay R Bay N Bay Bay J Bay A.3

FLAC (Version 4.00)

LEGEND

23-Apr-04 6:52 step 1986276 HISTORY PLOT Y-axis :X Displacement (Nd 29)X Displacement (Nd 628)X Displacement (Nd 629)X Displacement (Nd 392)X Displacement (Nd 568) X-axis :Dynamic time

5 10 15 20 25 30 35 40 45

-1.000

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References Beaty, M.H. and Byrne, P.M. (1999). A Synthesized

Approach for Modeling Liquefaction and Displacements; Proc. FLAC and Numerical Modeling in Geomechanics; Detournay and Hart (eds); 1999 Balkema, Rotterdam; pp. 339-347.

Youd, T.L., Hansen, C.M. & Bartlett, S.F. (1999).

Revised MLR equations for predicting lateral spread displacement. Proceedings of the Seventh U.S.-Japan Workshop on Earthquake Resistant Design of Lifeline Facilities and Countermeasures Against Soil Liquefaction, Technical Report MCEER-99-0019,

Multidisciplinary Center for Earthquake Engineering Research, State University of New York, Buffalo, N.Y.

* Driven further following PDA testing and prior to Statnamic Testing

** Ultimate Capacity is greater than the mobilized Capacity

*** Deflection at pile head (approximate only)


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