tunnel support- use of lattice girders in sedimentary rock

8/13/2019 Tunnel Support- Use of Lattice Girders in Sedimentary Rock

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Tunnel Support

Use of Lattice Girders in Sedimentary Rock

Einar Hrafn Hjlmarsson

Faculty of Civil and Enviromental

EngineeringUniversit of Iceland


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Tunnel support

Einar Hrafn Hjlmarsson

30 ECTS thesis submitted in partial fulfilment of aMagister Scientiarumdegree in Civil Engineering

SupervisorSigurur Erlingsson, University of Iceland

AdvisorHaukur Eirksson, Hnit engineering

Faculty RepresentativeGsli Eirksson

Faculty of Civil and Environmental EngineeringSchool of Engineering and Natural Sciences

University of Iceland

Reykjavk, October 2011


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Tunnel Support, use of lattice girders in sedimentary rock.Tunnel Support.30 ECTS thesis submitted in partial fulfilment of a

Magister Scientiarumdegree in Civil Engineering.Copyright 2011 Einar Hrafn HjlmarssonAll rights reserved

Faculty of Civil and Environmental EngineeringSchool of Engineering and natural SciencesUniversity of IcelandVR-II, Hjararhaga 2-6107, Reykjavk

Iceland

Telephone: +354 525 4000

Bibliographic information::Hjlmarsson, E.H., 2011, Tunnel support, use of lattice girders in sedimentary rock,Masters thesis, Faculty of Civil and Environmental Engineering, University of Iceland, 78

pages.

Printing: Hsklaprent ehf.

Reykjavk, Iceland, October 2011


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AbctractIcelandic geology is highly influenced by its location on the Mid Atlantic ridge. Changesin geological conditions are therefore frequent in Icelandic tunnels and flexible rocksupport methods are required to deal with constant variation of rock mass properties.

Main focus of this thesis is tunnel excavation and rock support in sedimentary rock mass.Instability problems in thick sedimentary rock layers in the shl tunnel are discussed.Installed support is evaluated using finite element modelling of the rock mass and bearingcapacity calculations of the tunnel lining. Special attention is given to the usage of latticegirders and how it was used as both temporary support to secure safer workingenvironment at the tunnel face and as a part of the final rock support for the tunnel.

Good estimation of rock mass properties to construct a reliable finite element model isimportant. Limited laboratory test data of the sedimentary rock mass turned out to be a

problematic factor. A number of uni-axial compression tests were made on core samplesfrom sedimentary layers in the tunnel but no tri-axial tests were executed. That led to veryconservative approach in the estimation of the rock mass parameters. Result from themodelling indicates that the tunnel lining used in the shl tunnel is sufficient if averagematerial parameters are used for the sedimentary rock mass. The rock support althoughseems to be slightly insufficient if lower limit (90% of test data above) is used for sedimentstrength. These results emphasise the importance of estimation of rock mass properties in

preparation of tunnel constructions.


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IndexFigures ................................................................................................................................. ix

Tables .................................................................................................................................. xii

Acknowledgement ............................................................................................................. xiii

1 Introduction ..................................................................................................................... 1

2 Rock mass properties...................................................................................................... 22.1 Icelandic rock mass ................................................................................................. 32.2 Characteristics of rock mass .................................................................................... 5

2.2.1 Generalized Hoek-Brown criterion ................................................................ 5

2.2.2 Mohr-Coulomb failure criterion .................................................................... 62.3 Stresses in rock mass ............................................................................................... 82.4 Discontinuties in the rockmass .............................................................................. 10

3 Types of rock support ................................................................................................... 113.1 Rock bolts .............................................................................................................. 11

3.1.1 General ......................................................................................................... 113.1.2 Types of rock bolts....................................................................................... 12

3.2 Shotcrete ................................................................................................................ 143.2.1 General ......................................................................................................... 143.2.2 Shotcrete methods ........................................................................................ 16

3.2.3 Shotcrete mix ............................................................................................... 163.2.4 Steel fiber ..................................................................................................... 173.2.5 Curing of shotcrete ....................................................................................... 17

3.3 Wire mesh .............................................................................................................. 183.4 Shotcrete ribs ......................................................................................................... 19

3.4.1 General ......................................................................................................... 193.4.2 Types of shotcrete rips ................................................................................. 203.4.3 Installation procedure (Norwegian standard)............................................... 223.4.4 Usage............................................................................................................ 22

3.5 Lattice girders ........................................................................................................ 233.5.1 General ......................................................................................................... 233.5.2 Types of lattice girders ................................................................................. 243.5.3 Installation procedure ................................................................................... 25

3.6 Steel arches ............................................................................................................ 29

4 Estimation of required rock support .......................................................................... 304.1 Rock quality designation index (RQD) ................................................................. 304.2 The RMR-system .................................................................................................. 314.3 Rock tunneling quality Index, Q-system ............................................................... 33

4.3.1 Determination of the Q-value ...................................................................... 334.3.2 Required rock support for estimated Q-value .............................................. 34

4.4 Numerical analysis ................................................................................................ 36


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5 Deformation of rock mass ............................................................................................. 375.1 Theory .................................................................................................................... 375.2 Modeling deformation curve .................................................................................. 39

5.2.1 Axisymmetrical FEM model ....................................................................... 40

6 Case study: Use of lattice girders in shl tunnel. .................................................... 436.1 Introduction ............................................................................................................ 436.2 Rock mass properties in shl tunnel .................................................................. 456.3 Estimation of rock mass parameters ...................................................................... 466.4 Calculated capacity of installed support ................................................................ 516.5 FEM-modelling ...................................................................................................... 54

6.5.1 Modelling method ........................................................................................ 556.6 Result ..................................................................................................................... 57

7 Conclusions .................................................................................................................... 62

Bibliography ....................................................................................................................... 63

Appendix ............................................................................................................................. 65A.1 Rock classification systems. ..................................................................................... 65A.2 GSI and D value in Hoek-Brown criterion ............................................................... 70A.3 shl tunnel: Geological mapping for st 16.300-16.450 ........................................ 72A.4 Bearing capacity of shotcrete lining. ........................................................................ 73


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FiguresFigure 2.1 Simplified geological map of Iceland (Weisenberger, 2010). ............................. 3

Figure 2.2 Mixed face. Basalt layers with approximately 2m sediment interbed. ................ 4

Figure 2.3 Mohr-Coulomb and Hoek-Brown failure criterions (Hoek, 2000) ...................... 7

Figure 2.4 H vs depth at various locations in Iceland. .......................................................... 9

Figure 2.5 Ideology of most stress criterias ......................................................................... 10

Figure 3.1 Fixing of a single block and systematic bolting(Palmstm et al, 2000) ........... 11

Figure 3.2 Most used rock bolts in Icelandic tunnels. Figures from (StatensVegvesen, 2000) ............................................................................................... 13

Figure 3.3 Application of spiling bolts in a weakness zone. ............................................... 14

Figure 3.4 Shotcreting ......................................................................................................... 15

Figure 3.5 Deflection curves for concrete slabs with and without steel fibers(Palmstm & Nilsen, 2000) .............................................................................. 17

Figure 3.6 Normal and measured curing of C30 shotcrete (Einarsson, 2010) .................... 18

Figure 3.7 Wire mesh to support local unstable area. ......................................................... 19

Figure 3.8 Radial rock bolting of a shotcrete rib (Vegagerin, 2008) ................................ 20

Figure 3.9 Single layered shotcrete rib (Statens vegvesen, 2009) ....................................... 21

Figure 3.10 Double layered shotcrete rib (Statens vegvesen, 2009) ................................... 21

Figure 3.11 Installation of single layer shotcrete ribs in Stjrdal tunnel, Norway(Gumundsson, 2011). ..................................................................................... 23

Figure 3.12 Thee and four-chord lattice girders (Jorimann, 2010) ..................................... 24

Figure 3.13 Scaling of under breaks and loose rock blocks ................................................ 25

Figure 3.14 Initial layer of shotcrete ................................................................................... 26

Figure 3.15 Installation of lattice girder .............................................................................. 26

Figure 3.16 Girder half embedded in shotcrete ................................................................... 27

Figure 3.17 Installation of spiling bolts ............................................................................... 27


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Figure 3.18 Lattice girder fully embedded with shotcrete .................................................. 28

Figure 3.19 Steel arch with sliding gap (Hoek et al.,2008) ................................................. 29

Figure 4.1 Example of RQD-value estimation (Hoek, Practical Rock Engineering,

2000) ................................................................................................................. 30

Figure 4.2 Rock support according to the Q-system (Hoek, 2000)..................................... 34

Figure 4.3 Generated element mesh of finite element model. ............................................ 36

Figure 5.1 Deformation vector around advancing tunnel (Hoek, Practical RockEngineering, 2000) ........................................................................................... 37

Figure 5.2 Longitudial deformation profile based on elastic and empirical models forRt =4,6m and Rp =7,45m. ................................................................................ 39

Figure 5.3 Axisymmetric model of the tunnel .................................................................... 40

Figure 5.4 Deformation vs decreasing inclusive e-modulus. .............................................. 41

Figure 5.5 Deformation curve for walls .............................................................................. 41

Figure 6.1 Overbreak due to weak sedimentary layer in shl tunnel. ............................. 43

Figure 6.2 Use of lattice girders and spiling bolts in shl tunnel. ................................... 44

Figure 6.3 Typical layered rock mass strata in the Vestfirir peninsula

(Gumundsson et al., 2007). ............................................................................ 45

Figure 6.4 Normal distribution of UCS data ....................................................................... 47

Figure 6.5 UCS vs Ei-module .............................................................................................. 47

Figure 6.6 Hoek and Brown failure envelop for lower limit strength parameters ofsediment............................................................................................................ 48

Figure 6.7 Hoek and Brown failure envelop for mean strength parameters ofsediment............................................................................................................ 48

Figure 6.8 Hoek and Brown failure envelope for basalt ..................................................... 50

Figure 6.9 Installed tunnel support at station 16.400 and 16.410. ...................................... 51

Figure 6.10 Cross section of lattice girder used in shl tunnel. Units are in mm. .......... 51

Figure 6.11 On left: Assembled lattice girders (Efla, 2011). On right: Final shotcretelining with lattice girders (Pedersen, Kompen, & Kveen, 2010). .................... 51

Figure 6.12 UCS result of shotcrete in shl tunnel. ........................................................ 53

Figure 6.13 Case 1 and 2. Station 16.400............................................................................ 54


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Figure 6.14 Case 3 and 4. Station 16.410. ........................................................................... 55

Figure 6.15 Maximum wall deformation and plastic radius. X indicates a shearfailure in the rock mass and o indicates tension failure. ................................... 55

Figure 6.16 Deformation curve for case 1 ........................................................................... 56

Figure 6.17 Inclusion modulus vs deformation ................................................................... 56

Figure 6.18 Result for case 1. Lower limit strength of sediment 12m in thicknessfrom tunnel invert. ............................................................................................ 57

Figure 6.19 Results for case 2. Mean strength of sediment 12m in thickness fromtunnel invert. ..................................................................................................... 58

Figure 6.20 Results for case 3. Lower limit strength of sediment 12m in thicknessfrom mid section. .............................................................................................. 58

Figure 6.21 Results for case 4. Mean strength of sediment 12m in thickness from midsection. .............................................................................................................. 59

Figure 6.22 Highest moments on the boarder of sediment and basalt in case 3. ................. 60

Figure 6.23 Result for case 4 with variable k value from 0,5 to 1,5. .................................. 61


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TablesTable 3.1 Properties of commonly used rock bolts (based on Palmstm & Nilsen,

2000) ................................................................................................................. 14

Table 3.2 Typical mix for C40 wet Shotcrete ..................................................................... 16

Table 4.1 Rock mass Rating system-RMR (enlarged in appendix A1) (Hoek,Practical Rock Engineering, 2000) ................................................................... 31

Table 4.2 Guidelines for rock support and excavation of 10 span tunnel according toRMR system. (Hoek, Practical Rock Engineering, 2000) ................................ 32

Table 4.3 ESR values (Hoek, Practical Rock Engineering, 2000) ...................................... 34

Table 4.4 Typical parameters for estimation of Q value in Iceland (Loftsson, 2009). ....... 35

Table 5.1 Rock mass parameters for axisymmetric FEM-model ........................................ 40

Table 6.1 UCS values of intact sedimentary rock samples in shl tunnel ...................... 46

Table 6.2 Rock mass properties of lower limit strength of sediment rock mass. ................ 49

Table 6.3 Rock mass properties of mean strength sediment rock mass. ............................. 49

Table 6.4 Min requirements for shotcrete in shl tunnel ................................................ 52

Table 6.5 Main model parameters ....................................................................................... 54


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Acknowledgement

I would like to thank the following people for their advice and assistance during my workon this thesis.

Dr. Sigurur Erlingsson for his guidance, helpful discussions and high interest in theproject.

Haukur Eirksson for his guidance and helpful advice.

Bjrn Hararson for providing data and literature.

Hlynur Gumundsson for providing data and pictures.

Gsli Eirksson for providing test data from shl tunnel.

Special thanks to my parents for all their support.


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1 IntroductionFirst road tunnel in Iceland was excavated through Arnadalshamar in 1948. Since then 9other road tunnels have been constructed along with numerous tunnels associated withconstruction of hydro power plants. Each tunnel has contributed to the knowledge of theIcelandic rock mass characteristics and usage of classification systems or other methods toestimate required rock support in the tunnels.

Tunnel support methods in Iceland have mainly been derived from experience andconventions in the Norwegian tunnelling industry and the Norwegian standard oftunnelling has been the foundation of Icelandic tunnel design. Q-value is used in the

Norwegian standard to classify rock mass into rock support categories. Norwegian rock

mass is however completely different from the Icelandic rock mass and adjustments hastherefore been needed to adopt the Q-system to Icelandic tunnelling conditions.

One of the newest challenges in Icelandic tunnelling was excavation through thicksedimentary layers in shl tunnel (also referred to as Bolungarvk tunnel). Rock supportmethod based on the usage of lattice girders was used for the first time in Icelandictunnelling to deal with low strength sedimentary rock layers.

Detailed description of the usage of lattice girders in sedimentary rock will be carried outin this report along with short overview of other main support methods used in Icelandictunnelling. Pros and cons of all methods will be evaluated.

Numerical analysis will be used to estimate deformation and stresses acting on such liningfor real cases from shl tunnel. Bearing capacity of reinforced shotcrete lining withlattice girders will be calculated according to the Eurocode 2 standard.

Foundation of high quality numerical analysis is the input parameters used to describe therock mass behaviour around the underground opening. Main properties of rock masscharacteristics will therefore be discussed and methods used to estimate rock mass

properties of sedimentary layer in shl tunnel.

Support method based on lattice girders has been used around the world for decadesbecause of their flexibility and easy assembling ability. Since the knowledge of the usageof lattice girders in Iceland is very limited a detailed description of their application inshl tunnel is considered contributing to the knowledge base of the Icelandic tunnellingindustry.


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2 Rock mass propertiesReliable design of an underground excavation requires reliable estimation of the rock masssurrounding the underground opening (Hoek, 2000).

Rock mass is a complex matrix of intact rock and weakness zones and number ofparameters influence its strength, deformability, permeability and stability behaviour.According to Kirkaldie (1988) at least 28 parameters could be used to describe rock mass

behaviour (Kirkaldie, 1988).

10 rock material properties

10 discontinuities properties

8 hydro geological properties

To include all these parameters in estimation of rock mass is hard or even impossible.

Limited number of the most representative parameters is therefore often used to describethe rock mass behaviour. Number of design and classification systems has been derived forthis purpose and some of them are described in later chapter (Palmstm & Nilsen, 2000).

Compression and shear strength of rock along with properties of the discontinuities in therock mass is usually the foundation of classification and design system in geotechnicalengineering. Knowledge of the virgin stress field around the underground opening is alsonecessary to estimate deformations and stress concentrations around the opening

The estimation of rock mass properties is usually the most challenging factor ingeotechnical design. Quality of the structure design is never higher than the quality of the

data used to estimate the rock mass parameters. Pre-geological investigation playstherefore an important role in the design process of every underground structure andshould be comprehensive enough to make the design believable and reliable.

Rock mass properties are highly dependent on its origin. Properties of igneous rock whichhave solidified from magma differ slightly dependent of the geological circumstance itsformed in but rock mass of sedimentary type has completely different characteristics. Basicknowledge of geology is there for needed to choose a suitable design or classificationsystems.


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2.1Icelandic rock massIceland is located on the boundary between the Eurasian and North American tectonic

plates called the Mid Atlantic ridge. The plates drift apart at the speed of 1-2 cm/yr and thevoid between the plates is constantly filled with igneous rock.

The North Atlantic opened around 60 million years ago and has from that time formed theNorth Atlantic sea floor generated from the modern day plate boundary the Mid Atlanticridge (Sigmundsson, 2006).

Figure 2.1 Simplified geological map of Iceland (Weisenberger, 2010).

The youngest rock is located near the volcanic active zones of Iceland and the oldest rock,14-16 million years old, is located in the north west and east coast. Icelandic bedrockconsists 80-90% of sub aerial basalt, 5-10% of acidic and igneous rocks and 5-10% ofsedimentary interbeds. Typical Icelandic bedrock consists of relatively thin basalt layerswith thin scoria layers on top and bottom of each layer and thin sedimentary interbeds. Thestrata usually dip slightly towards the volcanically active zone. Since the basalt layer areusually thin


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Figure 2.2 Mixed face. Basalt layers with approximately 2m sediment interbed.

Other rock formations like dykes and breccias are also frequently crossed duringexcavation of tunnel. Sedimentary intebeds is also frequent in Icelandic rock mass but areusually thin layered, from few centimetres to 1 or 2 meters. Occasionally thickersedimentary layers occur.

Main focus of this report is tunnel excavation in sedimentary rock and properties ofsedimentary rock will therefore be most discussed.

Sedimentary rock can be divided in two main categories by its formation:

1. Where material particles has been transported to the place of deposition, known as

clastic or detrital rock. Conglomerates, sandstones, siltstones, mudstones are of thatkind.

2. By an aggregation of organic matter or chemically/biochemically. Limestone, chalkand coal are of that kind.

Sedimentary rock is not interlocked like basalt rock but cemented together with anintergranual matrix by diagenesis (i.e. hardening of loose materials to rock) (Palmstm &

Nilsen, 2000).

Sedimentary rocks in the Tertiary bedrock are mostly fine grained tuffaceous interbeds andsome thicker conglomerates. The Pleistocene bedrock is usually more abundant of


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sedimentary rock, mostly sandstones and conglomerates of glacial origin (Hararsson,1991).

Thick sedimentary interbeds can cause some serious instability problems specially if joinedwith low strength, swelling potentials or sensitivity to water exposure.

2.2Characteristics of rock massEstimation of rock mass characteristics is required to design an underground excavation.Methods such as the generalized Hoek-Brown criterion and Mohr-Coulomb failurecriterion can be used to describe the characteristic behaviour of rock mass like strength anddeformations. Measured data from core samples are often used to estimate the properties ofintact rock (no weakness planes) and from that point through empirical approach toestimate the behaviour of the overall characteristics of the rock mass surrounding an

underground opening. Strength of intact rock sample is usually higher than the overallstrength of the rock mass and method are therefore needed to convert measured data fromcore samples to the rock mass(Hoek, 2000).

2.2.1 Generalized Hoek-Brown criterionHoek and Brown proposed in the 1980s a method to estimate the strength and propertiesof a jointed rock mass called Hoek-Brown failure criterion. The method is based onestimation of interlocking between rock blocks and shear conditions in the joints. Thismethod was derived to be used to estimate strength of jointed rock mass where rock blocksare small relative to the excavation considered. The method has been modified over the

years but the version introduced in this chapter is a modified version from 2002 (Hoek,Carranza, & al, 2002). The following series of equations represents the criteria:

Eq. 2-1

Eq. 2-2 Eq. 2-3 12 16 / / Eq. 2-4where and are the maximum and minimum effective principal stresses at failure,mbis the Hoek-Brown constant for the rock mass and miis the Hoek-Brown constant forthe intact rock samples, sand aare constants related to the rock mass characteristics andis the uniaxial compressive strength of intact rock sample.GSIis the geological strength index introduced by Hoek in 1994 to simplify the conversion

between the intact rock strength and the rock mass strength.


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A disturbance factorDis used to take into account the disturbance from blasting and stressrelaxations in the rock mass.

To estimate the value of mithe equation 2-1 is used with S=1 and a=0,5 and becomes:

1,

Eq. 2-5

A series of triaxial test on core samples can therefore be used to determine the value of miand

.Authors of the method recommends that series of at least five triaxial tests should

be used (Hoek, Carranza, & al, 2002).Authors also recommend that the range of should be equally distributed between zeroand 0,5 time the intact compression strength.

can be set to zero in equation 2-5 to reveal the uniaxial compression strength of therock mass and becomes:

Eq. 2-6

The tensile strength of the rock mass can be found in a similar way by setting equal tozero and the tensile strength becomes:

Eq. 2-7

Figure that can be used to estimate the value of GSI and D can be found in appendix A2.

2.2.2 Mohr-Coulomb failure criterionThe Hoek-Brown failure criteria is well suited for jointed or heavily jointed igneous rocktypes like basalt. But for other rock types like sedimentary or metamorphic rock the Mohr-Coulomb failure criterion can be a better choice (Hoek, 2000).

Estimation of shear strength can be made by the Mohr-Coulumb eqution:


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tan Eq. 2-8

where is the internal friction angle of the intact rock sample, Cis the cohesion and isthe normal stress acting on the plain of failure (Erlingsson, 2009).The Mohr-Coulomb equation can also be written as:

1 sin 1 sin 21sin tan Eq. 2-9

Uniaxial compression strength and tensile strength can be derived from equation 2-9 byputting and to zero respectively, thus 21 s i n Eq. 2-10

21sin Eq. 2-11

Figure 2.3 Mohr-Coulomb and Hoek-Brown failure criterions (Hoek, 2000)


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2.3 Stresses in rock massStresses within the rock mass are of great interest when designing an undergroundexcavation.

Compressive stresses are by convention positive and tensile stresses have negative prefix.Both high and low stresses can cause instability problems in underground excavation. Ifstresses exceed the compression strength of the rock mass it yields and low stress conditioncan cause rock blocks to slide due to low normal stress in the joints.

Stresses surrounding an underground opening are influenced mainly by the stress conditionprior to the excavation (virgin stresses) and the geometry of the opening. Number ofcomponents influences the direction and magnitude of the virgin stress field.

Main components influencing magnitude and direction of the stress field are (Palmstm &Nilsen, 2000):

Gravitational stresses are caused by gravitational force that pulls all material towards thecentre of the earth. Gravitational stresses increase with depth and 0,027 MPa/m is acommon number to calculate the vertical components of the gravitational stress field.Poisson ratio is often used to describe the relationship between the vertical and horizontalcomponent of the gravitation stresses in elastic materials. Poisson ration does not howevercorrespond very well with stress conditions in jointed rock mass.

Topographic stressesare caused by uneven ground surface. Uneven surface causes stressconcentrations and uneven stress field in the rock mass near the ground surface like in the

bottom of deep valleys and mountain slopes. The horizontal component is the overall stress

field can become the dominant factor near the surface in such cases.

Tectonic stresses are mainly caused by plate tectonic or the continental drift. Major faultsand folds are caused by tectonic stresses. Variation in tectonic stresses is high and that isone on the reasons why the horizontal component of virgin stresses are much more difficultto estimate than the vertical one.

Residual stresses. During the earlier geological stages of the rock mass it locks in someinternal stresses.

Vetical component can generally be calculated by equation 2-12.

0,027 Eq. 2-12The letter kis usually used to describe the ratio between horizontal and vertical stressesand the horizontal stress field becomes:

Eq. 2-13

Origin and behaviours of virgin stress field can be very complex and hard to estimate. The

vertical component of the stress field can easily be calculated with certain accuracy but the


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horizontal components will always be a rough estimation unless some rock stressmeasurement has been performed. Stress measurements are time consuming and expensive

but also very necessary to perform if a design of an underground excavation need relativelyaccurate estimation of the horizontal stress field.

Number of stress measurements has been done in Iceland, mostly related to construction ofhydroelectric power stations. Figure 2.4 shows overview of various stress measurements inIceland.

Figure 2.4 H vs depth at various locations in Iceland.

Teigbjarg, Sandfell and Blanda (Haimson, 1981)

Reyarfjrur (Haimson & Rummel, 1981)

Frskrsfjrur tunnel (Dahle, 2005)

Figure 2.4 shows that horizontal stresses at relatively shallow depth gives in most caseshigher k value than measurements at greater depth. This is not surprising since bothtopographic and tectonical stresses are responsible for higher portion of horizontal stressesat shallow depth.

Measurements of horizontal stress around the world also show that the value of ktends tobe high at shallow depth and decreases at depth (Hoek, 2000).

0

100

200

300

400

500

600

700

800

0 5 10 15 20 25 30

depth[m]

Hmax[Mpa]

Reydarfjordur

Teigsbjarg

Sandfell

Blanda

Frskrsfjrurtunnel

v(0,027xd)

H/v=0,5

H/v=1,5


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2.4 Discontinuties in the rock massDiscontinuities divide the rock mass into blocks of different scales. The blocks can becategorised by their scale as (Palmstm & Nilsen, 2000):

1. First order fault block defined by larger weakness zones or faults.2. Second order blocks formed by small weakness zones or seams.3. Third order blocks formed by normal joints4. Small discontinuities such ad bedding or schistosity partings.5. Small fragments or grains in the rock.

Discontinuities that form blocks that fall into category 5 are usually considered as part ofthe rock property and generally included in the strength characterisation of the rock.

Shear strength of discontinuities is an important factor in estimation of characteristic of

jointed rock mass. Numerous criterias and test methods have been derived to estimateshear strength of discontinuities (Hoek, 2000), but since the main focus in this report iscemented sedimentary rock, a detailed description of those criterias is outside the scope ofthis report.

Figure 2.7 shows the main ideology of shear stress criterias in discontinuities.

Figure 2.5 Ideology of most stress criterias


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3 Types of rock supportRock support is added to improve stability of underground opening. Pre-geologicalinvestigation are used to reveal the main characteristics of the rock mass surrounding theopening and a design is made to deal with various geological conditions in the tunnel.Geological conditions often change very rapidly on the tunnel route and that requireflexible support methods that can be quickly adjusted to the current circumstances(Palmstm & Nilsen, 2000).

Rock bolts and shotcrete are dominant support methods in Icelandic and Scandinaviantunnelling industry and has been for a long time. Shotcrete rips were first prescribed as anoptional support method in Hvalfjrur subsea tunnel in 1996 but were not used until inFrskrsfjrur tunnel in 2003 (Hararson, 2011). Steel arches are not commonly used

but have been used for example in Oddskars tunnel and hydro tunnel in the Krahnjkarproject (Loftson, 2011)

Main rock support methods will be described in this chapter and there pros and consdiscussed.

3.1Rock bolts3.1.1 GeneralDevelopment of rock bolts began in the 1920s and has since then become the mostdominant support method in underground construction (Luo, 1999)

Rock bolting is a flexible method very commonly used for rock support. Rock bolts arefrequently used as initial support at the tunnel face to obtain safe working conditions forthe crew and they also form part of the final rock support (Palmstm & Nilsen, 2000).After securing loose block at the excavation face by spot-bolting, a systematic bolting isoften used. Then a previously defined pattern of rock bolts are installed based on thegeological conditions. Geological mapping, Q-value or other methods are mainly used forthe design of the required pattern to secure the rock mass. A systematic bolting can beinstalled at the end of excavation or during excavation of the tunnel. Figure 3.1 showsspot-bolting and systematic bolting.

Figure 3.1 Fixing of a single block and systematic bolting(Palmstm et al, 2000)


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applicable in rock mass where high deformation can be expected since its stiffness can leadto failure in the rock bolt (Statens Vegvesen, 2000).

Figure 3.2 Most used rock bolts in Icelandic tunnels. Figures from (Statens Vegvesen,2000)

Swelling rock bolts can be used as initial rock support but are not qualified as permanentrock support by the Icelandic road authorities. Swelling bolts have direct contact with therock mass and therefore in great threat of corrosion. Installation procedure is however veryfast and therefore favourable under special circumstances.

Self drilling rock bolts are sometimes used in very weak or heavily jointed rock mass.Sometimes its very hard to keep the bore hole unblocked after drilling and therefore theinstallation of normal rock bolts are hard or impossible. That problem can be avoid withthe self drilling rock bolts since it doesnt require stable bore hole since it drills itself in.Self drilling rock bolts are however expensive because the drill bit at the end of the boltcannot be retained (Statens Vegvesen, 2000).

Rock bolts are sometimes installed ahead of excavation to support assumed unstable rockmass or rock mass with insufficient rock cover. Those rock bolts are called spiling boltsand are installed in a fan shaped pattern oriented 10-25relativly to the tunnel axis. Spiling

bolts are usually fully grouted with spacing of 30-80cm (Palmstm & Nilsen, 2000).

Figure 3.3 shows typical application of spiling bolts.


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Figure 3.3 Application of spiling bolts in a weakness zone.

Table 3.1 Properties of commonly used rock bolts (based on Palmstm & Nilsen, 2000)

Typeofrockbolt

Diameter[mm]

Yieldstrength[kN]

Failureload[kN]

Elongationatfailure

Elongationfor3mbolt[mm]

Standardboltlength

[m]

Roundsteelbar 20 6070 100 8% 240 0.86.0

Deformedbar 20 120 150 3% 90 0.86.0

25 220 250 1% 30 0.86.0

CTbolt 20 120 150 3% 90 1.56.0

22 200 250 2% 60 1.56.0

Hollowbolt 27 100 130 8% 240 2.06.0

3.2Shotcrete

3.2.1 GeneralShotcrete is a widely used method for tunneling support. This type of rock support isobtained by spraying concrete on the rock surface. Shotcrete for rock support has beenused for several decades and has become increasingly popular because of its favorable

properties together with high capacity and flexibility (Palmstm & Nilsen, 2000).


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Figure 3.4 Shotcreting

The main advantages of shotcrete as rock support are (based on Palmstm & Nilsen,

2000):

Ready to use on short notice (batching plant on site)

Framework is not needed

Applicable in uneven excavation profile

Easy to combine with other support methods

Flexible deformation properties.

The main disadvantages are: Low tension strength (can be improved by mixing fibers or reinforce with casting

steel)

Can collapse when applied on swelling rock types

Hard to apply on rock types with low cohesion (some types of sedimentary rock)

Hard to apply on rock with flowing water.

Shotcrete can be used in multiple variations as rock support:


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1. Ordinary shotcrete spayed in layers of up to about 60-100 mm thickness.2. Mesh reinforced shotcrete. This is produced by first spraying a layer of concrete

before installing a mesh with typically 5-6 mm diameter steel bars. Then a secondand sometimes more layers are applied to cover the mesh entirely.

3. Fiber reinforced shotcrete. Steel or plastic fibers are mixed with the wet concrete.

4. In combination with steel ribs, lattice girders or steel beams.Al these mentioned methods are typically used with combination of rock bolt to fastenloose block or to fasten the wire mesh or steel sets (Palmstm & Nilsen, 2000).

3.2.2 Shotcrete methodsWet mix shotcreting is now the dominant shotcrete method in modern construction oftunnels and has almost replaced the previous dry mix method. The difference betweenthese two methods are that the wet-mix method uses fully mixed concrete and blows it outof the nozzle of the shotcrete robot using compressed air.

The mixing water is however not added to the dry-mix until in the nozzle in the dry-mixmethod. That gives the operator the change to adjust the dosage of water duringshotcreting. The wet mix method is however more convenient for ordinary tunnelshotcreting since it has less rebound, higher capacity and gives better working conditions.

3.2.3 Shotcrete mixWhen producing concrete one uses a various types of recipes based on what properties ofthe concrete one wishes to get. Table 3.1 shows a typical mix for concrete used forshotcrete support in tunnelling in Norway (similar in Iceland).

Table 3.2 Typical mix for C40 wet Shotcrete

Portland cement 470 kgMicro-silica 8%Aggregates 0-8mm 1670 kgSuperplasticizer (BNS) 5 kgPlasticizer (lignosulphonate) 3,5 kgSteel fibers 50 kg

Accelerator (modified silicate) 5%

The amount of micro-silica and accelerator are calculated as percentage of cement. Micro-silica in the mix improves the strength properties of the shotcrete and makes it easier todistribute the steel fibers in the fresh shotcrete. It also reduces permeability and improvesfrost resistance (Palmstm & Nilsen, 2000).


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3.2.4 Steel fiberUnreinforced shotcrete is a brittle material that experiences cracks and displacement underrelatively low tensile stress compared to its compression strength. Steel fibers have beenused since the late 1950s to improve the concretes tensile properties.

By adding 1 vol % of steel fibers into a shotcrete mix can increase the load capacity of aC50 shotcrete slab by 85% (Palmstm & Nilsen, 2000). The usage of steel fibers has madeit possible to use a combination of rock bolts and fiber reinforced shotcrete instead of fullcast lining. A full cast lining is a very time consuming process and expensive so replacingthat with rock bolts and fiber shotcrete is very economically favourable.

Figure 3.5 Deflection curves for concrete slabs with and without steel fibers(Palmstm& Nilsen, 2000)

Plastic (polypropylene) fibers can also be used in increase the tensile strength of concreteand are often used for surface shotcrete for example as fire-protection for PE-watermembrane.

3.2.5 Curing of shotcreteTo increase the curing rate of shotcrete it is necessary to use so called accelerator. Itsmixed with the shotcrete in the nozzle and reacts with cement in the shotcrete. The dosageis usually 4-8% of cement weight in the mixture.

Usage of accelerator gives shotcrete initial curing to some degree so it doesnt fall of therock surface. The influence of the accelerator on curing and measurements of compressionstrength of core specimens of shotcrete from shl tunnel can be seen on figure 3.6. After15 hours the compression strength has reach 20 MPa. Shotcrete with no accelerator willreach this value after 4-5 days (Einarsson, 2010).


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Figure 3.6 Normal and measured curing of C30 shotcrete (Einarsson, 2010)

When estimating the bearing capacity of shotcrete its necessary to have information on thecuring rate of the shotcrete. Bearing capacity of a shotcrete lining is time dependant but thestresses form the tunnel wall is mainly dependent on the distance from the tunnel face.Right timing to activate a tunnel lining is therefore critical and is discussed further inrelations with ground reaction curves in Chapter 5.

3.3 Wire meshWire mesh is commonly used with combination of rock bolts and shotcrete. Use of fiberreinforced shotcrete has however reduced the need for wire mesh since installation of wiremesh is time consuming and is therefore only used where fiber reinforced shotcrete is notconsidered to be sufficient (Palmstm & Nilsen, 2000).

Wire mesh is also commonly used in combination with shotcrete ribs and lattice girders todistribute forces from the rock mass to the bearing elements.

0,1

1

10

100

0,1 1 10 100

UCS[MPa]

Curingtime[hours]6%accelerator 4%acceleratorNoaccelerator Measuredshltunnel6%Measuredshltunnel5%


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Shotcrete rips are built by fastening rebar steel along the tunnel profile with rock bolts andthen covered with shotcrete. Diameter of the rebars is usually 20mm (Norwegian standard)and number of rebars and spacing between rips is adjusted to the local condition andrequired bearing capacity.

Figure 3.8 Radial rock bolting of a shotcrete rib (Vegagerin, 2008)

3.4.2 Types of shotcrete ripsShotcrete rips can be either single or double layered as shown on Figure 3.7 and 3.8. Singlelayered shotcrete ribs are constructed by tying the rebars to a fastening rod fixed at the endof the rock bolts. Main disadvantage with single layered rips are their low momentcapacity. Rebars on top and bottom of a double layered rip however provide the rip withmuch larger moment capacity.


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Figure 3.9 Single layered shotcrete rib (Statens vegvesen, 2009)

Figure 3.10 Double layered shotcrete rib (Statens vegvesen, 2009)


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3.4.3 Installation procedure (Norwegian standard)Before installing the first (or only) layer of rebars, the rock surface must be smoothenedwith shotcrete. This layer is often 100-200mm thick fiber reinforced shotcrete (Palmstm& Nilsen, 2000).

After smothering of rock surface, the following procedure takes place:

Rock bolts installed with 1-1,5 m separation based on rock class (Norwegianstandard)

Fixing bars are fastened on all rock bolts.

Reinforcement bars (20mm) are fastened (welded or tied) to the fixing bars. The rib is shotcreted so that minimal shotcrete cover is 50mm (75mm for subsea

tunnels)

40mm spacing from reinforcement bars to the smoothening shotcrete is required beforeshotcreting the rip. Min 110mm spacing between parallel rebars is required. Point 2-4 inthe procedure is repeated to build a double layer shotcrete rip (Pedersen et al, 2010).

3.4.4 UsageShotcrete rip can be considered as a beam that experience load from the deforming rockmass. Because of the circular formation of the rip the load from the rock mass istransferred to axial stress in the rib. The rib can however also experience moment wherethe point load or uneven loads from the rock mass occur. Steel reinforcement is supposedto grant the rip with enough moment capacity to withstand those moments.

Shotcrete rips are also commonly used where overburden of the tunnel is low. Wherestresses and deformation are low can the ribs be installed relatively far from the face if theshort time stability is sufficient to provide the workers with safe working conditions at thetunnel face. Therefore they can be installing in greater number to gain productivity. Figure3.11 shows installed shotcrete ribs in the Stjrdal tunnel where overburden is only 2-4m(Gumundsson, 2011).


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Figure 3.11 Installation of single layer shotcrete ribs in Stjrdal tunnel, Norway(Gumundsson, 2011).

Main advantages of shotcrete rips are:

Material is on site or available with short notice.

Low preparation time.

Adapt to irregular profile.

High moment capacity if double layered

Main disadvantages are:

Needs to be rock bolted in the whole contour.

Difficult to use as a support for spiling bolt if profile is irregular.

Low moment capacity if single layered.

3.5Lattice girders3.5.1 GeneralLattice girders have been used for tunnel support since the late 1970s. In recent decade amove from heavy rolled steel arches to lighter, more manageable lattice girders has taken

place (Komselis et al, 2005).

Steel arches and lattice girders have basically the same function. They can be used as anelement of temporary support lining or a part of the permanent lining. Lattice girders werefor example used as part of the permanent lining in the underground transport system inBavaria, Munich in Germany with great success (Baumann & Betzle, 1984).


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Lattice girders are very similar to normal shotcrete ribs, the main thing distinguishing thesetwo support methods are the installation method. The final product is a bar reinforcedshotcrete lining.

Main advantages of lattice girders are (based on Komselis et al, 2005):

Simple and fast installation

Solid support for spiling bolts

Temporary support for shotcrete until in gains sufficient strength to support itself.

Immediate support in the area of the tunnel face.

High moment capacity

Main disadvantages are:

Needs to be ordered in time or constructed on site (min 1-2 weeks in Iceland).

Leads frequently to increased usage of shotcrete

3.5.2 Types of lattice girdersTwo types are most commonly used, three-chord and four-chord. Three-chord lattice hastriangular section with a larger bar (25-40mm) at the apex and two smaller diameter bars atthe base corners. Sinusoidal bars (10-12mm) separates the apex and the main bars. Four-chord lattice has four equally sized bars (20-40mm) at the corners of rectangular section.Sinusoidal side bars and cross bars (16mm) separates the main bars.

Figure 3.12 Thee and four-chord lattice girders (Jorimann, 2010)

Design diameters stated in this chapter are the most common used bar diameters in lattice

girder production. Other bar sizes is also possible (Komselis et al, 2005).


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3.5.3 Installation procedureInstallation procedure of lattice girders is highly related to the geological circumstancesand the purpose on their installation. It can be used as a temporary rock support to secure asafer working environment for the workers at the face or considered as part of the final

rock support lining.

Following description of installation procedure is considered applicable for installation oflattice girders where sedimentary rock is partly or the only rock type in the profile andwhere the rock mass stresses are not preventing that final support lining is at least partlyinstalled very close to the tunnel face.

The installation procedure is in ten steps and every step will be described.

1) Proper scaling of loose rock blocks is always the first step in every sequence ofinstallation of rock support. It minimizes the risk of fall down of shotcrete andmakes the working environment safer at the tunnel face. Scaling of under breaks inthe profile must also take place at this stage before shotcreting.Scaling in sedimentary rock with very low cohesion strength can though be risky.Few cubic meter of lost shotcrete is then better than risking large overbreaks byscaling to harshly with an excavator.

Figure 3.13 Scaling of under breaks and loose rock blocks


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2) After scaling the tunnel walls and roof must be secured by installation of shotcrete.Installation of the initial layer of shotcrete in sedimentary rock can be very time andmaterial consuming if the rock is weak. The weight of the shotcrete is sometimescausing too much shear stress in the rock and causing both shotcrete and rock tofall down. The shotcrete operator should though be able to build up the shotcreteshell by starting at the floor or harder rock and slowly cover the unsupported area.

Figure 3.14 Initial layer of shotcrete

3) In this case(as usually) the lattice girders is the first structural member to beinstalled since the initial shotcrete layer is not considered a structural member but asafety measure and is not involved in bearing capacity calculations since itsthickness varies.The lattice girder is assembled on the tunnel floor and raised by an excavator or by

the drill rig. To make sure that the girder is at the right position it must be check bya laser guidance system or a surveyor.

Figure 3.15 Installation of lattice girder

4) The girders must be fastened into place. All girders must at least be fastened asclose to the floor as possible. This is done to prevent that the end of the girderslides from the wall when forces from the deforming rock mass starts to act on the


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girder. Extra rock bolt can be added if hard stable rock is somewhere in the profile.

5) Wire mesh is usually installed behind the lattice girder. It both adds tensionalstrength to the shotcrete between the girders and also help with transferring the loadto the main bearing units which in this case is the bar reinforced concrete beam

build up by lattice girder and shotcrete. Rebars must be installed to overlap thejunctions of the girder to unify the moment and shear capacity of the girder.

6) Shotcrete is now used to cover the wire mesh and lattice girder. In this round thegirder is only half embedded in shotcrete to prevent that the girder is overloadedwith uncured shotcrete that is not able to support itself.

Figure 3.16 Girder half embedded in shotcrete

7) The girder is now used as a cantilever for spiling bolts used to support the rock fornext blasting round. The holes for spiling bolts are drilled through the lattice girderand the bolts installed. By installing a tight series of spiling bolt the rock ahead has

been supported and chances of overbreaks reduced.

Figure 3.17 Installation of spiling bolts


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8) The next tunnel round is now excavated.

9) After repeating stage 1-5 above in order to install the next lattice girder thepreviously installed girder is connected to the new one with side bars and coveredwith shotcrete. A continuous supporting lining is achieved by connecting all girderswith sidebars.

Figure 3.18 Lattice girder fully embedded with shotcrete

10)Installation of the previous lattice girder is now finished and stage 6-10 can now berepeated for the current lattice girder.

As previously stated is this working procedure of lattice girders adjusted for use inconditions where sedimentary rock layers are causing instability at the tunnel face. In other

geological condition the work procedure could be slightly different and must always beadapted to the local conditions. Similar work procedure was successfully used in theconstruction of shl tunnel.


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3.6Steel archesSteel arches and lattice girders are very similar construction elements in tunnels. They canwork as a temporary support or as a part of final support lining. Arches are often used as atemporary support in multiple drift tunnelling since it easy to disassemble (U.S.

Department of transportation, 2011).

Steel arched are rarely used in Iceland but has been used in Krahnjkar hydro project andOddskar tunnel.

Based on (U.S. Department of transportation, 2011) and (Hoek et al., 2008):

Main advantages of steel arches are:

Immediate support after installation

Excellent support for spiling bolts

Easy to disassemble and for forming temporary shotcrete wall in multiple driftexcavation

High deformability if granted with sliding joint. See Figure 3.19

main disadvantages are:

Heavy and expensive

Poor bonding with shotcrete

Figure 3.19 Steel arch with sliding gap (Hoek et al.,2008)


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4 Estimation of required rock supportEstimations of required rock support is usually based on observation, experience and

personal judgment of those involved. Engineers generally uses support guidelines ormethods to back up their estimation of required rock support. Three types of methods aremostly used for this purpose (Palmstm & Nilsen, 2000):

Analytical methods, involving analysis of stress distributions and deformationsusing methods like numerical analysis, analogue simulation or physical modeling.

Observational methods, like the New Australian Tunnelling Method which usesmeasurements of movements in the rock mass during excavation. Observations areoff course also used to check if the chosen installed rock support was the right wayto go or not and adjustment made if required.

Empirical methods, often illustrated in table or graphs that connect classification of

rock mass to curtain rock support. Number of empirical methods has been derivedsuch as the RMR system end the Q-system.

General overview of the most commonly used empirical methods will be given in thischapter and the use of the Final Element as an analytical method will be used in casestudies later on in this report.

4.1 Rock quality designation index (RQD)Deere et al. developed the RQD system in 1967 to estimate the quality of rock mass fromdrill cores. The RQD value is defined as the summarized length of all core pieces longer

than 100mm divided by the total length of drilled core (Hoek, 2000).

Figure 4.1 Example of RQD-value estimation (Hoek, 2000)

Nowadays the RQD system is mostly used as a sub-system in the RMR and Q-system.


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4.2The RMR-systemThe RMR-system was introduced by Bieniawski in 1973. The tunnel route is divided intosection where the geological properties are the same or similar and the tunnel support isassigned to those sections according to Table 4.1 (Palmstm & Nilsen, 2000).

Six geotechnical parameters are used to classify the rock mass (Erlingsson, 2009):

1. Uniaxial compression strength2. RQD-value3. Spacing of discontinuities4. Conditions of discontinuities5. Ground water conditions6. Orientation of discontinuities

Table 4.1 Rock mass Rating system-RMR (enlarged in appendix A1) (Hoek, 2000)


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Results from the RMR classification are given in number from 0-100. Table 4.2 showshow guidelines have been derived for rock support in 10m span tunnel for estimated RMRvalue.

Table 4.2 Guidelines for rock support and excavation of 10 span tunnel according toRMR system. (Hoek, 2000)


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4.3Rock tunneling quality Index, Q-system4.3.1 Determination of the Q-valueThe Q-system has been the dominated rock mass classification system since 1980 in

Iceland. The system was developed by Barton et al. in 1974 based on tunnelling experiencemainly in Norway and Sweden (Palmstm & Nilsen, 2000). In this classification systemare the joints and other discontinuities measured or estimated from core samples or visiblerock mass surface and thereafter the Q-value estimated for the rock mass. (Loftsson, 2009)

The Q-value is a number that varies on logarithmic scale from 0,001-1000 and is definedas:

Eq. 4-1

where:

RQD is the Rock Quality Designation is the joint set number is the joint roughness number is the joint alteration is the joint water reduction

The Q-value can be considered as a function of tree parameters; where each parameterrepresents different aspects of general rock mass strength (Erlingsson, 2009):

is and estimation of block sizes.

is an estimation of shear strength in joints.

is an estimation of the active stresses in the rock mass.

Table to determine the value of individual parameter of the Q-system can be found in theAppendix A1.

The Q-system has been used in Iceland to estimate required rock support in tunnellingsince 1980. Determination of individual parameters has been adjusted to Icelandic rockmass based on experience in various tunnel projects in Iceland (Loftsson, 2009). TheIcelandic road Administration published a report in 2009 (Loftsson, 2009), where thisexperience is used to establish some guidelines to estimation of individual parameters ofthe Q-values for different types of Icelandic rock mass. An overview of the guidelines fordetermination of different rock types for Icelandic rock mass is given in Table 4.4


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4.3.2 Required rock support for estimated Q-valueBarton et al. introduced in 1974 an addition parameter, Excavation Support Ration (ESR)to take into account a factor of safety correlated to the importance of tunnel construction.Table 4.3 gives Bartons suggested ESR values for different types of tunnel constructions.

Table 4.3 ESR values (Hoek, 2000)

Figure 4.2 shows estimated support needs for given Q-value and span of tunnel (Hoek,2000). Support giuedlines from the Norwaigan standard for estimated Q-value is in theAppendix A1.

Figure 4.2 Rock support according to the Q-system (Hoek, 2000).


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Table 4.4 Typical parameters for estimation of Q value in Iceland (Loftsson, 2009).

Rocktype RQD value Jnvalue

Basalt Usually 60-85 for low or medium jointed. 30-60for heavily jointed rock mass Usually 9 but 12 for flaky olivine basalt.

ScoriaFor hard well cemented, 50-80 but 30-50 for lowcemented rock mass

Usually no joint systems. 9-12 is used for well cemented and 15-20 for weak cemented.

SedimentValues of 40-60 are common but 10-30 if layered.If UCS 4. Up to 8-12 if filling are preventing interlocking ofrock blocks.

Dykes Usually 2-3 if joints are irregular. Usually 2-3. If filling include clay then 4.

Jw SRF

Basalt

1 if the tunnel is dry < 5l/min. 0,66 if waterpressure is < 2,5 bar and 0,33 if water pressure is >2,5 bars.

Usually 60-85 for low or medium jointed. 30-60 for heavilyjointed rock mass

Scoria


For hard well cemented, 50-80 but 30-50 for low cemented rockmass

Sediment


Values of 40-60 are common but 10-30 if layered. If UCS


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4.4 Numerical analysisNumerical modelling of rock mass is done by dividing the rock mass into large number ofindividual elements and use computer to calculate how they interact for given stress andstiffness conditions. The finite elements method used in this report is in the category of

continuous models which means that the rock mass is considered to be continuous rockmass and only a limited number of discontinuities (joints, faults, etc.) may be used. Toconstruct a finite element model of rock mass one must define geological properties of therock mass and the virgin stress field. Element mesh of the rock mass is then generated todivide the rock mass into individual elements and boundary conditions determined. Finallythe magnitude of stresses and deformations are calculated for every nodal point in the meshfor given rock mass and boundary conditions (Palmstm & Nilsen, 2000).

The computer program Phase2 (version 7) is used in this report to construct finite elementmodel of rock mass surrounding tunnel in sedimentary rock mass. The FEM is used tocalculate the deformations occurring during excavation and stresses that reacts on installed

rock support. Phase2 is able to automatically generate the element mesh. An example ofgenerated mash can be seen on Figure 4.3. Notice how the elements get smaller as it closeson the excavation profile.

Figure 4.3 Generated element mesh of finite element model.


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5 Deformation of rock massStability in advancing tunnel in weak rock is strongly related to the deformation occurringafter excavation. Deformations of the rock mass starts about one and half diameter of thetunnel in front of the tunnel face and has reached 25-33% of the final deformation alreadyat the tunnel face. In weak rock mass these deformations might already result in instabilityat the tunnel face depending on the rock mass strength and the in situ stress conditions(Hoek, 2000).

Figure 5.1 Deformation vector around advancing tunnel (Hoek, Practical RockEngineering, 2000)

In order to determine the appropriated timing of the installation of specific tunnel supportone must have knowledge of the rock mass deformation behaviour and displacementcapacity of the support (Hoek et al., 2008).

Creation of longitudinal deformation profiles and ground reaction curves are convenienttools to gain knowledge of the rock mass behaviour. Longitudinal deformation profiles can

be measured in situ or created by using analytical or numerical methods such as the finiteelement method (FEM). Use of analytical approach is however only possible for very

simple shape of excavation (circular) and isotropic elastic or elastic-plastic rock mass. Tomodel more complex shaped excavation profile and rock mass with plastic behaviour onemust use axisymmetric or three dimensional finite element models. Empirical best fitanalysis can though be used to create deformation profiles if maximum displacement and

plastic radius has been measured or modelled.

5.1TheoryPanet derived in 1995 a relationship base on elastic analysis for short term longitudinaldeformation profile for known maximum deformation umax(Hoek et al., 2008):


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14

34 1

33 4

Eq. 5-1

where dt=X/Rt,Xis the distance from face andRtis the runnel radius.

Alternative expressions have been derived for elastic analysis for deformation profilessuggesting that the deformation doesnt form a continuous curve in front and behind theadvancing tunnel face. Unlu and Gercek suggested in 2003:

1 0 Eq. 5-2

1

0 Eq. 5-3

where u0is the deformation at the tunnel face andAa, Ab, BaandBbare functions of thePoisson ratio:

0,22 0,19 Eq. 5-4 0,22 0,19 0,73 0,81 Eq. 5-5

0,22 0,81 0,39 0,65 Eq. 5-6where is the Poissons ratio.

Numerous empirical best fit solutions based on plastic modelling and empirical best fit toactual measured closure date have been suggested.

Based on measured data, Chern in 1998 suggested the following expression (Hoek, er al,2008):

1 , , Eq. 5-7

The shape of the deformation curve is however directly connected to the radius of theplastic zoneRp. Therefore Hoek et al. (2008) have suggested the following relationship forshort term longitudinal deformation curve:

13 , Eq. 5-8

where Pris given as:

Eq. 5-9


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0 Eq. 5-10

1 1

0 Eq. 5-11Relationship for short term longitudinal deformation curve derived by Hoek et al.(2008)will be used as part of the case study in Chapter 6. Figure 5.2 shows comparison of elasticand empirical models mentioned above.

Figure 5.2 Longitudial deformation profile based on elastic and empirical models forRt =4,6m and Rp =7,45m.

5.2Modeling deformation curveAn axisymmetric FEM analysis can used to model the response of an elastic-plastic rockmass around non-circular cross section. The process can best be explained by an example.In this chapter, a axisymmetric model and best fit empirical analysis is used to construct adeformation curve for a given rock mass and field stress circumstances.

Cross section and rock mass properties in Table 5.1 from shl tunnel are used in thisexample. Rock mass properties are analyzed data from Mannvit consulting engineeringoffice from core samples taken during construction of the tunnel (Mannvit, 2009).

0

0,1

0,2

0,3

0,4

0,5

0,6

0,7

0,8

0,9

1

20 10 0 10 20 30 40

Ur/Umax

Distancefromface[m]

Hoek

Chern

Elasticmodel

Rt=4,6m

Rp=7,45m


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Table 5.1 Rock mass parameters for axisymmetric FEM-model

Rocktype Stiffness

Tensilestrength

Frictionangle Cohesion UCS

HoekBr.parameter

HoekBr.parameter

HoekBr.parameter

Em T0 c ci mb s a

[MPa] [MPa] [] [MPa] [MPa]

Basalt 12270 80 4,09 0,00855 0,503

Sediment 2266 0,79 56 1,66

The field stress is considered to be 16,2 MPa in vertical direction and 6,48 MPa inhorizontal direction. That corresponds to 600m over burden and k value of 0,4. Thisexample should be considered as hypothetical and will not be used in the case study inChapter 6.

5.2.1Axisymmetrical FEM model

For this example a tunnel of 9,2 m width is considered to be crossing a 12m thicksedimentary rock embedded in a basalt rock mass. Location of the sedimentary layer isconsidered to be in the invert of the tunnel cross section.

Figure 5.3 Axisymmetric model of the tunnel

As the tunnel advances the support from the tunnel face is reduced. To simulate this effectan elastic material is placed inside the excavation profile and its stiffness reduced in stages.At stage 1 the E-modulus of the inclusive rock mass is the same as the surrounding rockmass and zero at the final stage to simulate a fully excavated tunnel far from the tunnelface. Deformations for a single point can therefore be collected for each modulus of theinclusion material. Figure 5.4 shows the results of the deformations in the walls in theexample.


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Figure 5.4 Deformation vs decreasing inclusive e-modulus.

According to the axisymmetric model (see Figure 5.4) the maximum average mid walldisplacement is 74mm for unsupported tunnel far from the tunnel face. The maximum

plastic radius can be estimated at the final stage of the model as 7,45m.

Deformation curve as a function of distance from the tunnel face can now be created byusing the empirical relationship suggested by Hoek (see Chapert 5.1) since both maximumdeformations and plastic radius have been estimated in the model. The curve is given inFigure 5.5.

Figure 5.5 Deformation curve for walls

The deformation curve also shows how much of the total displacement has already taken

0

250

500

750

1000

1250

1500

1750

2000

2250

2500

0 0,02 0,04 0,06 0,08 0,1

InclusiveE

module

[MPa]

Walldeformation[m]

Leftwall

Rightwall

Average

0

0,01

0,02

0,03

0,04

0,05

0,06

0,07

0,08

10 0 10 20 30

Walldeformation[m]

Distancefromface[m]

Deformationcurve

InstallingofsupportInstallingofsupport1,5mfromface


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6 Case study: Use of lattice girders inshl tunnel.

6.1Introduction

shl tunnel is located in the Vestfirir peninsula at the north west corner of Iceland andconnects two small towns, Hnfsdalur and Bolungarvk. The tunnel is a 9,2m span roadtunnel with 53,75 m2cross section and a length of 5,2 km. shl tunnel were constructedin 2008-2010 and are categorized as low traffic tunnel according to the Norwegian roadtunnel standard with annual average daily traffic (AADT) around 780. The tunnel wasexcavated by traditional drill and blast method.

This case study involves a FEM-design of a shotcrete lining through sedimentary rock inthe shl tunnel. Pre-geological investigation indicated number of sedimentary interbeds

along the tunnel route. Original design of the tunnel support suggested that thosesedimentary interbeds would be dealt with by conventional rock support like rock bolts,shotcrete, wire mesh and shotcrete ribs with spiling bolts according to convention inIcelandic tunnel construction.

During the construction of the tunnel some instability problems were experienced in thetunnel crown when excavated through sedimentary interbeds that in some cases lead tocollapse of the tunnel crown right after blasting (Figure 6.1). An alternative workingmethod for tunnel support was therefore initiated to deal with these circumstances. Thatsupport method involved installation of lattice girders and spiling bolts.

Figure 6.1 Overbreak due to weak sedimentary layer in shl tunnel.


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Since the use of lattice girders was not expected during the design of the tunnel, the tunnellining involving lattice girders was designed on site with participation of engineers from

both contractor and the supervision and approved by the tunnel designer.

Figure 6.2 Use of lattice girders and spiling bolts in shl tunnel.

The FEM program, Phase 2, is used to estimate the forces that react on the tunnel liningunder the current geological circumstances. The design code Eurocode 2 is further used toestimate the bearing capacity of the lining.

Only limited investigations of the rock mass properties were available. Number of uni-axial compression tests (UCS) were made to estimated the intact rock strength of bothsedimentary interbeds and basalt. No tri-axial tests results were available to estimate thegeological parameters in the Hoek-Brown or Mohr-Coloumb failure criterion. To

investigate how much this limitation of available data will infect the design both averageand lower limit strength parameters were used.

At station 16.360 (TM 2.095 from Hnfsdalur side) a sedimentary layer began to arise fromthe bottom of the tunnel. 10m thick sedimentary interbed had been logged in core hole atst. 16.100 and the current layer was considered to be the same one. Exploratory holesindicated that the layer could be 12-15m thick and according to the inclination of the stratait would be in the profile for at least 80-100m. Excavation revealed though that tectonicfault had shifted the strata in station 16.390 and excavation in sediment layer turned out to

be around 50m. Geological circumstances around station 16.400 to 16410 is used in thiscase study since UCS and cohesion test data are available for that sedimentary layer.

Geological mapping can be found in the Appendix A3 of the area discussed.


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6.2Rock mass properties in shl tunnelshl tunnel are located in the Vestfirir peninsula at the north west corner of Iceland.The rock mass of the tunnel route is part of the oldest rock mass in Iceland. The rock massstrata is formed by highly alterated basalt layers and numerous sedimentary layers, some

with very low compression strength. The rock mass strata dips 5-7to south west and sincethe tunnel route is almost parallel to the dip direction it goes through around 500m of therock mass strata. Most basalt layers are around 3-8m thick but goes up to 20m inthickness. (Gumundsson et al., 2007)

Pre-geological investigation of the tunnel route indicated that the tunnel route would crossnumerous sedimentary layers with thickness from 1-5m. Much thicker sedimentary layerswas though encountered during excavation of tunnel and lead to use of heavier rocksupport than previously anticipated.

Figure 6.3 shows typical layered rock mass strata in the Vestfirir peninsula.

Figure 6.3 Typical layered rock mass strata in the Vestfirir peninsula (Gumundsson

et al., 2007).


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Figure 6.4 Normal distribution of UCS data

Figure 6.5 UCS vs Ei-module

Measurements of cohesion strength in this same borehole indicated mean cohesion strengthof c= 0,79MPa and c10% = 0,37 MPa.

The stiffness modulusEi10% should be around 1366 MPa according to Figure 6.5.

Other UCS measurements of sedimentary rock in shl tunnel indicates UCS values from1,5-9,8 MPa (Plsson, 2009). That supports the conservative approach of using UCS10% .

0

0,01

0,02

0,03

0,04

0,050,06

0,07

0,08

0,09

0,1

0 2 4 6 8 10 12 14 16 18 20 22 24 26 28 30Probability

Density

UCS [MPa]

Density

10% 90%

0

500

10001500

2000

2500

3000

3500

4000

4,0 6,0 8,0 10,0 12,0 14,0 16,0 18,0 20,0StiffnessE

[MPa]

UCS [MPa]


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Figure 6.6 Hoek and Brown failure envelop for lower limit strength parameters ofsediment.

Figure 6.6 shows the Hoek-Brown failure envelope (red) for the lower limit strength of thesediment layer. Figure 6.7 shows the Hoek-Brown failure envelope if mean values are usedfor UCSstrength and correspondingEi-modulus.

Figure 6.7 Hoek and Brown failure envelop for mean strength parameters of sediment.


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Figure 6.8 Hoek and Brown failure envelope for basalt


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6.4Calculated capacity of installed supportInstalled tunnel support in all cases that will be analysed (see further in Table 6.5) arelattice girders c/c 1500mm, 150mm minimum fiber reinforced shotcrete cover betweengirders and 6mm wire mesh behind the girders. Figure 6.9 shows cross section of the

installed tunnel lining and Figure 6.10 shows cross section of the lattice girder.

Figure 6.9 Installed tunnel support at station 16.400 and 16.410.

Figure 6.10 Cross section of lattice girder used in shl tunnel. Units are in mm.

Figure 6.11 On left: Assembled lattice girders (Efla, 2011). On right: Final shotcretelining with lattice girders (Pedersen, Kompen, & Kveen, 2010).


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UCS-testing on core samples displayed in Figure 6.12 shows that 99% of core samples hadhigher UCS strength than 30MPa.

Figure 6.12 UCS result of shotcrete in shl tunnel.

Material partial factor for concrete c is given as 1,5 in Eurocode 2. Design strength ofshotcrete in shl tunnel should therefore be (Eurocode 2, 2002):

30

1,5 20 Eq. 6-1Eurocode 2 however notes that the partial factor c can be lowered if reliable measurementscan be used to confirm the real strength of the concrete. Partial factor s for steel is 1,15.

0

1

2

3

4

5

6

7

8

910

29 30 31 32 33 34 35 36 37 38 39 40 41 42 43 44 45 46 47 48 49 50 51 52 53 54 55 56

Numberoftests

UCSstrengthofshotcrete[MPa]


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6.5 FEM-modellingFour cases will be modelled that represents the geological circumstances at stations 16.400and 16.410. Geological mapping of the area can be found in Appendix 4. Estimated rockmass parameters from Chapter 6.3 are here used to construct FEM-models that are

considered to have similar characteristic as the rock mass in those locations in shltunnel. Table 6.5 shows the main model parameters for each case.

Table 6.5 Main model parameters

Case1 Case2

Location st16.400 Location st16.400

Sedimentlayer 12m Sedimentlayer 12m

Positionofsediment Frominvertup Positionofsediment Frominvertup

Rockmassparameter Lowstrength(Table6.2) Rockmassparameter eanstrength(Table6.3)

Virginstressfield 400moverburden,K=0,5 Virginstressfield 400moverburden,K=0,5

Case3 Case4

Location st16.410 Location st16.410

Sedimentlayer 12m Sedimentlayer 12m

Positionofsediment Frommiddlesectionup Positionofsediment Frommiddlesectionup

Rockmassparameter Lowstrength(Table6.2) Rockmassparameter eanstrength(Table6.3)

Virginstressfield 400moverburden,K=0,5 Virginstressfield 400moverburden,K=0,5

Rock mass parameters for the basalt are given in Figure 6.8

Figure 6.13 and 6.14 shows position of the sedimentary layer (brown) in the tunnel profileeach case.

Figure 6.13 Case 1 and 2. Station 16.400.


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Figure 6.14 Case 3 and 4. Station 16.410.

6.5.1 Modelling methodThis chapter shows how axisymmetrical modelling of deformation curves are used toconstruct reliable FEM-model of the geological situations of each case. All cases aremodelled in the same way and modelling process of case 1 is therefore only shown in thischapter,

Case 1.St: 16.400, 12m Sedimentary layer from tunnel invert with lower limit strength.

Figure 6.15 Maximum wall deformation and plastic radius. X indicates a shear failurein the rock mass and o indicates tension failure.

The analysis shows that maximum wall deformation for unsupported tunnel far from the

tunnel face is 40mm (80mm convergence) and maximum plastic radius is 7,26m. Figure


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Figure 6.17 created with same method as described in Chapter 5.2, shows that the tunnelface is still supporting the tunnel profile 1,5m from the face with similar effect as ifmaterial with stiffness of 500 MPa where placed inside the excavation.

Tunnel relaxation is therefore generated until the inclusion modulus is 500 MPa. It

corresponds to 1,5m from the tunnel face. All deformation after that will build up stressesin the tunnel lining.

All four cases are modelled by the same method but different inclusion modulus accordingto their corresponding axisymmetric deformation model.

tunnel support- use of lattice girders in sedimentary rock

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