1 INTRODUCTION

Flow slides associated with heavy rainfalls are often

devastating, causing many casualties in many parts

of the world. However, their development has not

been well understood until recently, in particular be-

cause the stress paths followed in slopes subjected to

pore pressure increase were not considered and the

concepts of instability of cohesionless soils were not

well understood. When a slope is subjected to pore

pressure increase due to infiltration or rising water

table, total stresses and shear stresses remain essen-

tially constant but effective stresses, mean effective

stress in particular, decrease. This corresponds to a

specific stress path which, examined within the con-

cepts of critical state and instability, provides a gen-

eral framework for understanding slope behaviour in

both loose and dense soil deposits (Chu et al. 2003).

The first part of the paper focuses on the onset of

slope instability (i.e. development of plastic strains),

and post-failure as such is not considered. In the

second part, evidences from physical models and

case histories supporting the framework are pre-

sented. Finally, there is a discussion on the practical

implications of the framework on the evaluation of

slope stability.

2 MECHANISMS LEADING TO SLOPE INSTA-

BILITY

Before considering slopes, it is useful to examine

some aspects of cohesionless soil behaviour. When a

soil is consolidated, isotropically or anisotropically,

in a triaxial cell and then subjected to an undrained

compression test, the stress-strain curve shows a

peak and then moves towards an ultimate state, often

called steady state or critical state. Figure 1 presents

such stress paths in a q/p’cs vs p’/p’cs diagram, in

which p’cs is the mean effective stress at the critical

state. Sladen et al. (1985b) called the line joining the

peaks obtained in CIU tests to the critical state (CS)

the Collapse surface. At a given void ratio and in a

p’ vs q diagram, that would be a collapse line; at a

different void ratio, the collapse line is different.

Lade (1993) defines these lines as instability lines

(IL). The zone bounded by the IL and the critical

state line (CSL) is the zone of instability in which

loose sand becomes unstable when an undrained

condition is imposed (Chu et al., 2003; Wanatowski

et al., 2009a). The IL also appears to be a state

boundary surface that is followed by soil elements

that reach it to move towards the critical state. This

was evidenced by Sasitharan et al. (1993) who per-

formed constant shear drained (CSD) tests, with

constant deviatoric stress and decreasing mean effec-

tive stress from initial stress conditions at point I, on

loose Ottawa sand (Fig. 2). The specimen collapsed

at point Y, well below the (CSL), at a void ratio of

0.809. The collapse (or instability) line correspond-

ing to the same void ratio is also shown on the fig-

ure. It can be seen that yielding in the CSD test has

been obtained when the stress path reached the in-

stability line, indicating that instability is associated

with the same effective stresses and void ratio, re-

gardless of drainage conditions.

Slope instability due to pore water pressure increase

S. Leroueil Laval University, Quebec City, Quebec, Canada

J. Chu Nanyang Technological University, Singapore

D. Wanatowski The University of Nottingham, Nottingham, United Kingdom

ABSTRACT: Chu et al. (2003) developed a framework based on specific stress paths followed in slopes sub-

jected to pore water pressure increase and concepts of soil instability and critical state. This framework applies

to loose as well as dense sands. In this paper, these concepts are slightly extended and confirmed on the basis

of physical model observations and case studies. Practical implications related to the evaluation of slope sta-

bility and the significance of calculated factor of safety are then discussed.

From Sladen et al. (1985b)’s data (see Fig. 1), it

turns out that, if stress conditions on the collapse

surface are described by an angle of strength mobili-

zation φ’mob (e.g. corresponding to M6.9 for the test

consolidated at 6.9 p’cs), this angle is lower than the

critical state friction angle φ’cs and increases towards

its value when the consolidation stress p’/p’cs de-

creases towards one.

Been & Jefferies (1985) proposed the state pa-

rameter Ψ for characterising the behaviour of sands.

Ψ is defined in a e vs p’ diagram (Fig. 3) as the dif-

ference between the current void ratio eo (at a point

such as I) and the void ratio on the steady state line

(or CSL), ess(I), under the same mean effective stress,

Ψ = eo – ess. Positive values of Ψ are associated with

contractant soil behaviour whereas negative values

of Ψ are associated with dilatant soil behaviour.

In the context of slopes subjected to pore water

pressure increase, the stress paths followed are at an

essentially constant shear stress with decreasing ef-

fective stresses. In the laboratory, such stress paths

can be simulated by drained tests with constant q

value and decreasing p’ value called constant shear

drained (CSD) tests (Brand, 1981; Anderson & Rie-

mer, 1995). Such tests had been performed by

Anderson & Riemer (1995) on colluvial soil and by

Santos et al. (1996) on residual soil. On the basis of

the results, Leroueil (2001) indicated that instability

is not controlled by initial stress conditions at point I

(Fig. 3) and the corresponding Ψ value but rather by

the relative position of the point at yielding Y (Fig.

3) with respect to the CSL. Chu et al. (2003) sug-

gested calling this difference the modified state pa-

rameter __

Ψ (Fig. 3). In fact, the value of Ψ in a slope

varies with seasons whereas slope instability and

possibly failure depends on __

Ψ .

Chu et al. (2003) performed a series of triaxial

tests, including CSD tests, on Changi sand having a

mean grain size of 0.30-0.35 mm and a uniformity

coefficient of 2.0. In the interpretation of the tests,

Chu et al. (2003) defined the term instability as be-

haviour in which large plastic strains are generated

rapidly due to the inability of a soil element to sus-

tain a given stress or load. Figure 4 illustrates insta-

bility under CSD tests. Two specimens of loose sand

(ec = 0.94), DR7 and DR10, were anisotropically

consolidated at p’ = 200 kPa (point A for DR7, Fig.

4a) and then sheared at an essentially constant devia-

toric stress at a rate not allowing the development of

pore pressure during the test that was thus fully

drained. From A to B, there were little axial and

volumetric strains (Fig. 4b). However, at point B,

these strains started developing at a faster rate, indi-

cating unstable behaviour. As shown in Figure 4a,

the instability point B is well below the critical state

line, at an angle of strength mobilization φ’mob lower

than φ’cs.

Two specimens of the same sand, but dense (ec =

0.65), DR39 and DR40, were anisotropically con-

solidated at about p’ = 260 kPa (point D for DR39,

Fig. 4a) and then sheared at an essentially constant

deviatoric stress. As shown on Figure 4b, from D to

E, the axial and volumetric strains were small. How-

ever, at point E, the axial strain started increasing,

indicating the onset of instability; as shown in Fig-

ures 4b and 4c, this latter instability was associated

with dilation of the dense soil specimens. It can be

seen that in these cases, instability is reached at an

angle of strength mobilization φ’mob larger than φ’cs.

Figure 4c shows the location of the instability

points for the previously mentioned tests in an e vs

log p’ diagram. The critical state line is also shown

on the figure. It can be seen that the modified state

parameter __

Ψ is equal to about 0.060 for the loose

specimens and about -0.186 for the dense specimens.

From these tests and other triaxial tests performed

on the same sand, Chu et al. (2003) defined the rela-

tionship between the effective stress ratio at instabil-

ity, ηIL = q/p’, and __

Ψ (Fig. 5a). It can be seen that

ηIL, equal to 1.35 at critical state, varies from about

0.70 for __

Ψ > 0.10 to about 1.50 for __

Ψ < - 0.15.

Wanatowski & Chu (2007) and Wanatowski et al.

(2009b) performed plane strain tests on the same

Changi sand and observed a general behaviour simi-

lar to that obtained in triaxial conditions. However,

the critical state lines obtained in both tests were dif-

ferent with, in particular, M values equal to 1.35 for

triaxial tests and 1.16 for plane strain tests. The rela-

tionships between the effective stress ratio at insta-

bility, ηIL, and __

Ψ were also different. However,

Wanatowski et al. (2009a) showed that, when nor-

malised with respect to the M value obtained in the

relevant type of test, triaxial or plane strain, both

curves come on a unique one (Fig. 5b).

From these results, Chu et al. (2003) developed a

general framework for understanding the instability

of slopes in loose or dense sand subjected to pore

water pressure increase. Leroueil (2004) summarised

it by using a q/p’cs vs p’/p’cs diagram in which the

CSL and IL (linear for simplicity) were drawn (Fig.

6). C is the normalised critical state, at the intersec-

tion of the CSL and IL. For normalised q values lar-

ger than the one at C (generally relatively loose

sands for slopes of precarious stability), initial con-

ditions will be at a point such as Ils in Figure 6 and

instability will be reached at a point such as Yls, be-

low the CSL. At Yls, the soil will have a tendency to

move towards its critical state C. As the deviatoric

stress at C is smaller than that due to gravity forces

in the slope (q at Ils), there will be static liquefaction

of the soil and collapse of the slope. Major conse-

quences of this phenomenon are that failure is trig-

gered at an angle of strength mobilization smaller

than the critical state friction angle and that instabil-

ity (at Yls) is followed by an increase in pore water

pressure since p’ decreases. If soil in a slope is at Ids,

at a normalised q value smaller than the one at C

(generally relatively dense sands), and is subjected to

pore water pressure increase, the stress path will

move towards Yds where there is development of

plastic strains and then have a tendency to go to-

wards its critical state C. However, this can only be

achieved if the soil dilates and q/p’cs increases,

which takes time.

This framework can be described in more details

by referring to Figure 7. When a loose sand is

sheared along a q = cst path starting from point I

(Fig. 7a), instability occurs at point Y, on the IL at

the corresponding __

Ψ value. If the pore water pressure

can dissipate freely (i.e. under drained condition),

the stress path will eventually reach the failure state

at point C1, on the critical state line that is also the

failure line for loose sands. During this process,

large axial and volumetric strains will develop and

the void ratio of the soil will decrease (Fig. 7b). If

failure is reached at C1, as there is then equilibrium

between the applied shear stress and the critical state

strength, the available kinetic energy will be small

and the rate of movement should be small (Leroueil

et al., 1996). On the other hand, if the pore-water

pressure cannot dissipate freely, the stress path will

move towards the critical state associated with its

current water content, i.e. C2 under undrained condi-

tions. As the corresponding strength is smaller than

the applied shear stress, the kinetic energy available

at yielding will be large. There will thus be runaway

failure and development of flow.

For dense sand, the stress path moves from point I

to point Y, where the soil becomes unstable (Fig.

7b). For dense sand, instability is reached above the

CSL but below the failure line. If the pore water

pressure can dissipate freely (i.e. under drained con-

ditions), both axial strain and volumetric strain (dila-

tive in that case) will start increasing at point Y, and

the stress state will move towards the failure line at a

point such as F1. If failure is reached, post-failure

could be essentially undrained, with a stress path

moving towards point C4 in Figure 7b. As the corre-

sponding strength may be smaller than the shear

stress applied by gravity forces in the slope, there

may be development of flow at the post-failure

stage. If at point Y, dilation cannot be accommo-

dated, such as for undrained conditions, negative

pore water pressure will develop (see Wanatowski et

al., 2009a) and the stress state will move towards

point C3 (Fig. 7b). The soil will then remain stable;

it could however become unstable and the slope

could possibly fail when pore water pressure will

dissipate. In such a case, failure is delayed.

Figure 7c shows a particular case in which insta-

bility is reached at the same time as the CSL, at

point C5. In that case, instability is associated with

failure. However, as there is then equilibrium be-

tween the applied shear stress and the critical state

strength, the available kinetic energy will be small

and the rate of movement should also be small.

3 EVIDENCE FROM PHYSICAL MODEL TESTS

AND FIELD OBSERVATIONS

The framework previously described has important

implications for slopes: In loose sand, the onset of

failure can be obtained at an angle of strength mobi-

lization, φ’mob, smaller than φ’cs; in such case, and if

essentially undrained the onset of failure is followed

by an increase in pore pressure. In dense sand, the

onset of instability is reached at a mobilized friction

angle, φ’mob, larger than φ’cs; however, instability is

then associated with a tendency of the soil to dilate

before failure can be reached.

That framework has been established for satu-

rated conditions. It is more complex for unsaturated

soils since, with infiltration, matric suction (ua – uw)

decreases, (p – uw) decreases and the strength enve-

lope is lowered. In addition, there does not seem to

have detailed information on what happens to insta-

bility lines with changing suction. However, and as

indicated by several reported case histories, the con-

cepts previously described apply. Olivares &

Damiano (2007) specify that when the soil is suscep-

tible to static liquefaction and is essentially saturated

at the onset of slope failure, post-failure will evolve

into a flow slide as for saturated soils. These authors

also mention that if the soil is not saturated at the

onset of failure, then post-failure may not evolve

into a flow slide but rather into a slide or debris ava-

lanche, with smaller runout distance.

As suggested by the US National Research Coun-

cil (NRC 1985), failure may also result from re-

distribution of void ratio within a globally undrained

sand layer or spreading of excess pore water pressure

in a slope. These latter possibilities are not examined

here. Also, it is tried to avoid in this paper cases

where failure could be associated with erosion or ex-

cavation at the toe of slopes. Such cases involve dif-

ferent stress paths and possibly a quasi-undrained or

partly drained behaviour (e.g. failure at the Jamuna

Bridge, Bangladesh, (Hight et al., 1999) and failure

of the Mississippi riverbanks (Torrey & Weaver,

1984)).

Several cases from the literature are examined

hereunder in comparison with the implications of

this framework.

Nerlerk berm. The Nerlerk berm was constructed

over two seasons in the Beaufort Sea, where the

depth of water was approximately 45 m. The berm

consisted of Ukaler sand core overlaid by Nerlerk

sand. The slope angle of the berm was approxi-

mately 13°. At the turn of July and August 1983,

several slides occurred, involving only Nerlerk sand.

This sand was assumed to have a relative density of

30% (Lade, 1993) and a friction angle at critical state

of 31°. Back analysing the failures, Sladen et al.

(1985b) found a mobilized angle of strength of 13-

16°, indicating instability considerably below the

CSL at a friction angle of 31°.

Coking coal stockpile physical model. Eckersley

(1990) examined flow slides in coking coal stock-

piles. The coal particles ranged from fine sand and

silt sizes to gravel with a critical state friction angle

of 40°. Instability was induced in 1 m high stock-

piles by raising the water level within the slope (Fig.

8a). For the experiment considered here (Experiment

7), the coal was placed for the bottom 400 mm at 9%

water content with no compaction (dry density of 0.7

Mg/m3); the remaining 600 mm was placed dry at a

dry density of 1.0 Mg/m3.

Failure occurred in three stages as indicated in

Figure 8a. Stage 1 comprised two shallow slides

over a 2 s period. It was followed by Stages 2 and 3

that occurred in the 4 following seconds along the

shear zones shown in Figure 8a. Pore water pres-

sures observed during experiment 7 at locations in-

dicated in Figure 8a are shown in Figure 8b. These

pore water pressures became positive during raising

of the water table and were slowly increasing at the

time of failure. What can be seen is that pore water

pressures mostly increased after the onset of failure

defined on the basis of video camera pictures (ar-

rows in Fig. 8b). Eckersley (1990) concluded: “Ex-

cess pore pressures are a consequence of failure ini-

tiation rather than a cause, and static liquefaction is

therefore a post-failure phenomenon”. He also back

calculated a mobilized angle of strength at the onset

of failure of 24-27°, much less than the critical fric-

tion angle of 40° obtained from laboratory tests.

Centrifuge tests. Zhang & Ng (2003; also reported

by Ng, 2008) have performed centrifuge tests for ex-

amining the failure mechanisms of sandy slopes sub-

jected to rainfall and rising water table. The material

used was Leighton Buzzard fine sand that shows

pronounced strain-softening in undrained triaxial

shear tests performed on loose specimens. The

model was 305 mm high with a slope of 29.4° and

built with the soil at a relative compaction of 68%.

However, when the model was subjected to an ac-

celeration of 60 g, the slope was densified to 80% of

the maximum relative compaction and flattened to

24°. At 60 g, the slope was de-stabilised by rising

water level and the soil liquefied statically and

flowed. Unfortunately, the pore water pressures re-

ported by Ng (2008) are not very detailed.

Flume tests. Wang & Sassa (2001) and Damiano

(2003) examined rainfall-induced flow slides in

laboratory flume tests. The two materials tested were

silica sandy silt and pyroclastic sand respectively.

Failure was induced by sprinkling water on the sur-

face of the soil models. Figures 9a and b show typi-

cal results obtained on these materials in loose con-

ditions. It can be seen in both cases that pore water

pressures slowly increased before the onset of failure

and rapidly increased after, similarly to Eckersley

(1990) observation (Fig. 8b).

Wachusett Dam. The construction of the dam was

completed in 1907 and failure occurred in the up-

stream slope during the first reservoir filling. Ac-

cording to Olson et al. (2000), the upstream fill con-

sisted primarily of fine sands that were placed

without compaction. Failure occurred in drained

conditions as filling of the reservoir was very slow

and was followed by a flow of the upstream fill soils

over a distance of about 100 m into the reservoir.

Back analysis of the failure performed by Olson et

al. (2000) indicate that failure occurred at average

shear strength between 37.6 and 41.9 kPa, corre-

sponding to a mobilized angle of strength that was

close to the critical state friction angle of the sand

(30°). Kinetics analysis also indicated a post-failure

strength of approximately 16 kPa, slightly less than

half of the strength at failure. A practical conclusion

from Olson et al. (2000) is that sandy fills that sub-

sequently will be saturated should not be placed

without compaction.

Cernivara landslide. The Cervinara landslide oc-

curred on 16 December 1999 along a steep slope of

about 40° covered by about 2.5 m of pyroclastic soils

and developed into a flow slide that travelled over

several kilometres, stroke several houses and killed 5

persons. The volcanic soils include layers of pumice

and volcanic ash classified as sand. Detailed studies

of the mechanical behaviour of volcanic ash have

been performed at the Seconda Università di Napoli.

Olivares & Picarelli (2001, 2003) showed that: (a)

strength significantly increases with matric suction;

(b) critical friction angle is of 38°; (c) the material is

highly susceptible to liquefaction when saturated

(see Fig. 10).

Suction measurements made on the same site give

values of 20 to 50 kPa during the dry season and 4 to

8 kPa during the wet season. The stability of the

slope was thus insured by matric suction. However,

it is thought by Olivares & Picarelli (2003) that on

16 December 1999 slope failed because suction van-

ished. The mobilized angle of strength was then

probably very close to the critical state friction angle

of the soil. In addition, as volcanic ash is very sus-

ceptible to liquefaction (Fig. 10), the landslide

turned into a flow slide.

Sau Mau Ping landslides. Two landslides involving

man-made fill slopes occurred after heavy rainfall on

18 June 1972 and 25 August 1976 at Sau Mau Ping

in Hong Kong (Ho & Sun, 2009). For the 1972 land-

slide, the slope was 40 m high and inclined at 34°

with the horizontal, and the debris slid down at high

velocity, killing 71 persons; the 1976 landslide took

place in a 35 m high and 33° steep slope, and turned

into mud flow, killing 18 persons. The slopes were

made up of decomposed granite. Investigation of the

1976 landslide showed that the fill was extremely

loose (ρd = 1.35 Mg/m3, corresponding to about 75%

standard compaction) to a depth of at least 2 m be-

low the slope; beyond the crest, ρd was also low and

variable, between 1.65 and 1.2 Mg/m3 (90% and

70% relative compaction), to a depth of 7 m and

about 1.5 Mg/m3 down to 20 m. Laboratory direct

shear box tests carried out under a vertical consoli-

dation stress of 25 kPa showed a contractant behav-

iour for dry densities lower than 1.5 Mg/m3; other

laboratory tests indicated a critical state friction an-

gle of 36.8°.

It was estimated that the fill could have been satu-

rated to depths between 2 m and 6 m under the 25

August 1976 rainstorm. Also, numerical analyses

showed that failure could have been triggered if 3 m

of loose fill became saturated, but for strength condi-

tions below the critical state strength envelope. The

recommendation following this investigation was

that the soil has to be compacted to not less than

95% of standard maximum dry density for man-

made fills (Ho & Sun, 2009).

Evidences of dilatant behaviours. The possibility of

dilatant behaviour of soil masses prior to some post

failure movements is also supported by observations:

(a) Casagrande (1975) indicates that prior to lique-

faction and flow of large masses of rather dense

granular talus in the alps, brooks emerging from the

toe of the talus stopped flowing; (b) Fleming et al.

(1989) report observations of time lags between the

beginning of landslide movements and the initiation

of debris flows; (c) in three slides that Harp et al.

(1990) triggered by artificial subsurface irrigation,

they observed abrupt decreases in pore pressure 5 to

50 minutes before failure.

4 DISCUSSION

The cases presented in the previous section confirm

the soil model and its implications for slopes. In

loose sandy soils, the onset of failure can be reached

at an angle of strength mobilization lower than the

critical state friction angle (Nerlerk berm; Coking

coal stockpile; probably Sau Mau Ping slope) or

close to the critical state friction angle (Wachusett

dam; Cervinara slope); where measured (and if the

soil is not too pervious), instability is followed by

pore water pressure increase (coking coal stockpile;

flume tests); in all cases, there was strain-softening

of the soil and development of post-failure flow

slide; in all cases also, development of flow slide is

very rapid. In the cases of dense soils, there is evi-

dence of dilatant behaviour and delay between the

onset of instability (development of plastic strains)

and failure as such. For the two slope cases reported

here that were generally unsaturated (Cervinara and

Sau Mau Ping), it seems that failure was initiated

when the matric suction was close to zero. It can be

thought however that some slopes failure can be

reached when they are still unsaturated (in loess in

particular); Olivares & Damiano (2007) indicate

however that in these conditions, the possibility to

have a flow slide is smaller.

4.1 Practical application of the framework

The framework and its implications being accepted,

the two main practical questions are as follows:

Considering instability, what are the mobilizable

strength parameters that should be considered in sta-

bility analyses of a slope subjected to pore water

pressure increase? What is the representativeness of

the factor of safety calculated by effective stress

analyses? Even if extremely important, post-failure

behaviour is not considered here.

A major practical question concerns the location

of the instability line relative to the critical state line

or, in other words, what is the angle of strength mo-

bilization to be considered. The answer is given by a

diagram such as Figure 5a and the ratio between ηIL

at relatively large __

Ψ values and M. It seems how-

ever that the ηIL/M vs __

Ψ relationship is soil specific.

ηIL/M at large __

Ψ values is about 0.5 for Changi sand

(Fig. 5b), about 0.55 for Leighton Buzzard sand (Fig.

1) and about 0.9 for Cervinara volcanic ash (Fig.

10). Testing 3 different Japanese sands, Orense et al.

(2004) concluded that instability was reached at mo-

bilized friction angle φ’mob such that tan φ’mob = 0.73

to 0.83 tan φ’cs. ηIL/M thus appears to be variable

from soil to soil and also with__

Ψ . Also, for the time

being, there is not enough data available to be able to

correlate ηIL/M with some physical characteristics of

the soils.

A simplified approach of the problem does not

seem to be accessible for the time being. On the

other hand, it appears difficult to develop a rational

and practical methodology for the approach pro-

posed by Chu et al. (2003) and described here. There

are several reasons for that: (a) the number of tests

that is necessary for determining the CSL and the ILs

at different void ratios is important; (b) natural soil

variability may also be a difficulty; (c) if reconsti-

tuted soil is used, the mode of preparation of the

specimens may significantly influence the test results

(e.g see Vaid et al., 1995). It is thus recommended,

when possible, to take undisturbed samples, recon-

solidate them under stresses close to in situ stresses

and then shear them at constant deviatoric stress and

decreasing mean effective stress, in CSD tests. It is

thought that the test results are the best indicators of

the behaviour of a slope under increasing pore water

pressure and can be directly used. For comparison, it

is also suggested to define φ’cs for the considered

soil.

4.2 Representativeness of a calculated factor of

safety

Another practical aspect is the representativeness of

a calculated factor of safety for a given slope. As

previously indicated and as illustrated by the stress

paths IY in Figs. 2, 4 and 7a, the initiation of failure

may be obtained for an angle of strength mobiliza-

tion smaller than φ’cs in loose sandy soil. This phe-

nomenon is amplified by conventional limit equilib-

rium stability analyses. In these analyses, the factor

of safety is calculated by comparing the applied

shear stress to the shear stress at failure under the

same normal effective stress, which implicitly as-

sumes an effective stress path such as IG in Figure

11 (see Tavenas et al., 1980), thus very different

from the stress path leading to failure, i.e. IY. The

calculated factor of safety for a slope in loose sandy

soil may thus significantly overestimate the real sta-

bility; for the case schematised in Figure 11, that

would mean a calculated factor of safety of about 2

(from I to G) whereas the slope is in fact close to

collapse (from I to Y).

4.3 Solutions to decrease the possibility of instability

and flow slides

As suggested by Olson et al. (2000) and Ho & Sun

(2009), a solution for decreasing the possibility of

soil instability and liquefaction is to increase the an-

gle of strength mobilization by compaction. Ng

(2008) and other authors suggest reinforcement of

the slope by methods such as nailing.

5 CONCLUSION

The framework established by Chu et al. (2003) and

re-examined here provides a unified way to study the

instability, failure and post-failure mechanisms of

loose and dense sandy slopes. For loose cohesionless

soils, instability is reached at an angle of strength

mobilization that is smaller than the critical state

friction angle. If perfectly drained, failure is reached

at the critical state with a rate of movement that

should be small. If not perfectly drained, soil insta-

bility is followed by pore water pressure increase,

flow and runaway failure. For dense soils, instability

(development of plastic strains) is reached at an an-

gle of strength mobilization slightly larger than the

critical state friction angle. If perfectly drained and

dilation allowed, failure is reached on the failure line

and may be followed by runaway post-failure. If di-

lation cannot be accommodated, negative pore pres-

sures develop and the slope will not fail.

Observations made in physical model tests and in-

terpretations of case histories confirm this frame-

work. The practical application of these concepts is

however difficult to apply and it is suggested, when

possible, to take undisturbed soil samples, recon-

solidate them under in situ stresses and then shear

them in CSD tests in order to evaluate their behav-

iour. It is also shown that calculated factor of safety

of slopes in loose sand can strongly overestimate real

stability.

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Géotechnique, 35(2): 99-112.

Brand, E.W. 1981. Some thoughts on rain-induced slope fail-

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