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STRUCTURAL PERFORMANCE AND ECONOMICS OF TALL

HIGH STRENGTH RC BUILDINGS IN SEISMIC REGIONS

BENEDICT T. LAOGAN AND AMR S. ELNASHAI*Department of Civil and Environmental Engineering, Imperial College, London, U.K.

SUMMARY

For a multitude of economic and societal considerations, high rise structures are on the increase. This in turnpromotes the use of high strength materials to reduce column size and construction times. Whereas designguidance and engineering understanding of high strength RC structures under static loading is well-developed,little work has been undertaken on the economics of whole buildings and their performance under earthquakeloading. In this paper, 10 buildings of 24 stories are designed and detailed according to modern seismic codes. Thebuildings are all nominally equivalent, using a stiffness equivalence criterion and its derivatives. The cost of

construction is compared in terms of steel, concrete and formwork. The static inelastic response of the buildings isalso assessed, followed by a full nonlinear dynamic analysis of all buildings using three earthquake records at thedesign acceleration and twice the design value. Comprehensive assessment of the static and dynamic results isundertaken. It is concluded that the cost increase is mainly due to the steel, whilst significant member reductionsmay be availed of by using high strength concrete. The behaviour of high strength concrete structures is notinferior to that of normal strength materials. Indeed, it is observed that lower levels of overstrength can beachieved in high strength materials than in their normal strength counterparts, mainly due to the over-reinforcement of the latter to resist vertical forces. Recommendations on the use of equivalent cracked stiffness forperiod calculation in design, and also effective periods for use in displacement-based design, are given. Copyright 1999 John Wiley & Sons, Ltd.

1. INTRODUCTION

1.1. High rise buildings

The increasing reliance of employment on financial services is one of the reasons that lead to

increasing rural-to-urban migration which in turn lead to increased demand on land use in large cities.

Whereas in 1950 there were only 7 urban areas in the world with more than 5 million inhabitants, this

number rose to 34 in 1980 and is expected to rise further to 60 by the year 2000.1 Consequently, more

high rise structures are being constructed now than a decade or two ago. A secondary stimulus for

construction of high rise buildings is that of an engineering challenge, whereby the two targets of

boasting the longest bridge and the highest building have become serious considerations in the

conceptual design of landmark projects. Examples of such projects are the Akashi-Kaikyo Bridge

linking Kobe City to Awaji Island (length 3910m) and the Petronas Towers in Kuala Lumpur (height

450m). The historical development of building height is shown in Figure 1.

The need for higher buildings naturally leads to the conclusion that high strength constructionmaterials will be increasingly used in the future, in order to keep column sizes at manageable

THE STRUCTURAL DESIGN OF TALL BUILDINGSStruct. Design Tall Build. 8, 171204 (1999)

CCC 10628002/99/03017134 $17.50

Copyright 1999 John Wiley & Sons, Ltd.

Received November 1998

Accepted April 1999

* Correspondence to: Professor Amr. S. Elnashai, Department of Civil and Environmental Engineering, Imperial College,Imperial College Road, London, SW7 2BU, U.K.

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dimensions and to make more effective use of floor areas, especially in the lower storeys of high rise

structures. Two other performance criteria lend weight to the argument for the use of high strength

concrete (HSC). Increased wind and traffic vibration susceptibility dictates that the modulus of

elasticity of the material should be as high as possible in order to limit small amplitude elastic

displacements. Moreover, the need for rapid construction requires early age strength gain, a feature

that may be offered readily by high strength concrete. The combined effect of the three above-

mentioned requirements renders high strength concrete economics rather appealing.

Modern seismic design codes require the designer to identify sources of energy absorption and

quantify their energy absorption capacity in a response range through which no collapse will occur.Therefore, the fact that new construction materials may offer higher strength to the designer than

conventional ones should not be a comforting thought. It is their deformation characteristics that

should be carefully studied from an earthquake resistance viewpoint. Indeed, there is a real danger that

static concepts, where additional strength implies additional safety, would lead to an unquantifiable

increase in seismic risk for high rise buildings. The only means of controlling this risk is the

development of comprehensive seismic design guidance for high strength materials to at least mirror

the state of development of codes for conventional materials.

1.2. History of high strength concrete and steel in construction

A continuous increase in the compressive strength of concrete has been observed in the past

decades, though with restricted availability. Today, high strength concrete is already being used for

high rise buildings, long-span bridges and offshore structures in many parts of the world.2 The

definition of high strength concrete has changed over time. In the 1950s, 35MPa concrete was

considered high-strength. By the 1960s, compressive strengths of 50MPa could be attained. This limit

increased to 60MPa in the early 1970s, whilst recent developments in concrete technology make it

possible for compressive strengths of up to 100MPa to be attained with relative ease. 3 In research

laboratories, concrete of strengths as high as 800MPa have been reported.4 The definition of high

strength concrete also varies with geographical location. In general, concrete of strength greater than

Figure 1. Historical evolution of building height

172 B. T. LAOGAN AND A. S. ELNASHAI

Copyright 1999 John Wiley & Sons, Ltd. Struct. Design Tall Build. 8, 171204 (1999)

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40MPa is considered high strength. However, in regions where 60MPa concrete is used extensively,

strengths of around 80 to 140MPa are considered high strength.2

An outstanding example of highly developed city centres resorting to high rise structures, using high

strength materials, is Chicago, Illinois, where the development of the use of high strength concrete

followed the trend indicated in Table I.

There are several outstanding examples of high rise buildings using high strength materials around

the world, both existing and under construction, or under consideration. In Seattle, Washington, the

Pacific First Centre (44 stories) and the Two Union Square (62 stories) employ concrete with

compressive strength of 115MPa, mainly used for its high modulus of elasticity of 50 000MPa. These

are extreme examples, and more commonly used high strength concrete mixes are represented by the

Two Prudential Plaza (281m) and 311 South Wacker Drive (295m), which use a mix with

compressive strength of 83MPa. Another structure is under considerationthe Miglin-Beiter building

with a height of 610m and concrete of 97 MPa compressive strength. Outside the U.S.A., the Petronas

Towers (450m) employ concrete of compressive strength 80 MPa for the lower stories, whilst the BfG

building in Frankfurt (186m) goes up to 85MPa.

Development of high yield steel for general construction lagged behind that of concrete. Only in the

early 1990s has high yield steel been widely available mainly from Japanese steel manufacturers, with

yield strengths above 1000MPa being tested and used.5 This has had an effect on high strength

materials construction economics, because the decrease in column dimensions would lead to anincrease in required steel area unless high yield steel is used.

1.3. Response characteristics of high strength concrete and steel

The main economic advantages of high strength concrete are its higher strength per unit cost,

strength per unit weight and stiffness per unit cost. The unit price of concrete increases relatively less

than the increase in available strength. This gives high strength concrete an economic edge in terms of

strength per unit cost. Moreover, the unit weight of concrete only increases slightly with increasing

compressive strength. This gives significant advantage in seismic areas, since earthquake forces are

directly proportional to the weight of the structure. Additional stiffness due to the increase in modulus

of elasticity leads to higher stiffness per unit weight and per unit cost for high strength concrete.6

Tests on the material level have indicated that unconfined high strength concrete has a near-linear

ascending branch and a steep linear descending branch, as indicated in Figure 2. This is mainly

attributed to the high strength of the matrixaggregate interface zone; thus, gradual breakdown of the

bond between them, to which the nonlinearity of the ascending branch is attributed, does not take

place. Another feature of significance to seismic design is the increase in strain at ultimate stress

relative to the increase in ultimate stress. In Figure 2, the stress ratio (high strength to normal strength

concrete) is about 24, whilst the corresponding axial strain ratio is about 17.With regard to the lateral stressstrain response, it is clear from Figure 2 that the dilation

Table I. Development of concrete strength in Chicago

Year Building name fc (MPa)

1962 Outer Drive East 411965 Lake Point Tower 52

1972 Mid-Continental Plaza 621976 River Plaza 761982 Chicago Mercantile 961988 Construction Tech. Labs. 117

HIGH STRENGTH RC BUILDINGS 173


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characteristics of the two materials are in stark contrast. This may be a consequence of the much-

reduced internal micro-cracking for high strength concrete, due to the similarity of the deformation

characteristics of the matrix and the aggregates. Indeed, the onset of unstable fracture propagation,

used to explain ordinary concrete response at high stress and the sudden increase in Poisson ratio, is

not noticeable for high strength concrete. In the transverse direction, the strain ratio is, from Figure 2,

nearly unity, contrasting with the stress ratio of 24 mentioned before. The implications for seismicperformance are rather grim. The material does not dilate appreciably, hence confinement is not

mobilized until a later stage in the stress history, and the loss in strength is sudden. It is therefore

possible that, all section characteristics being equivalent, an HSC section will have a lower curvature

ductility than the normal strength counterpart. However, there are doubts about this, arising from two

additional features of high strength concrete behaviour. Firstly, the level of axial stress on an HSCmember is significantly higher than in normal strength concrete. Hence, even if the dilation ratio is

less, the absolute value of lateral strain may be comparable. Secondly, stressstrain relationships for

high strength confined concrete are still under development, and the effectiveness of confining steel is

still very much debatable.

For under-reinforced members, the first clear departure from linearity of the momentcurvature plot

is close to the yield of tensile steel. Therefore, steel stressstrain behaviour is central to the evaluation

of ductility capacities. Most existing high yield steel reinforcing bars have no distinct yield point, as

shown in Figure 3. Also, in most cases, the ratio between ultimate and yield stress (stress ratio), for an

implied yield point of HYS, is lower for the latter than for conventional bars.

There are two implications of the above observations from a seismic design point of view. Firstly,

allocating a yield value to HSC/HYS sections is even more arbitrary than for conventional materials.

Secondly, as a consequence of the low stress ratio, plastic hinge lengths for high strength materials will

tend to be shorter than otherwise, thus deflection ductility will be lower for the same level of curvature

ductility. This conclusion is in contrast to several experimental studies in which it was reported that

HSC/HYS members can achieve ductility values comparable to conventional reinforced concrete (e.g.

Morita and Shiohara,5 Razvi and Saatcioglu7). In the latter studies, the conclusion was based on tests

of members under axial loads, or eccentric loads. Whilst such studies are valuable, they are not strictly

applicable to the earthquake response of buildings, where the axial load on columns does not

Figure 2. Typical stressstrain relationships for unconfined concrete



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necessarily increase in sympathy with the transverse load. Hence, only tests on members subjected to

constant axial load and increased, and varying, transverse displacement-controlled loading furnish

conclusions relevant to seismic design. Therefore, in the authors opinion, adequate ductility of high

strength materials, comparable to conventional ones, is by no means a settled issue, mainly because of

the role played by the extent of the plastic hinge (localization of inelastic deformations).

1.4. Code design considerations

There are concerns that the provisions of the existing codes cannot be applied to the design of high

strength concrete structures due to differences in material properties. Most of the empirical equations

available to designers are based on tests with normal strength concrete. The ACI Committee 363 State-of-the-Art Report on High Strength Concrete2 includes a chapter discussing the effect of high strength

concrete properties on the behaviour and design of structures at the member level. Comparison and

comments are made regarding applicability of existing ACI 3188 code provision for design using high

strength concrete. A summary of some of the key points and recommendations raised in the report is

given hereafter.

For columns under pure axial loading, the current design practice of adding directly the concrete and

steel strength is still valid for high strength concrete. Moreover, the use of the 085 factor in the designequation for the calculation of nominal axial load capacity was also shown to be satisfactory. On the

effectiveness of lateral confinement reinforcement, it was concluded that present equations are

applicable from the point of view of strength. However, the post peak stressstrain properties of high

strength columns designed based on these equations might be deficient.

Eccentrically loaded columns designed using a rectangular stress block compared to those designed

for the more realistic trapezoidal stress block show little difference. Furthermore, the moment

magnification method for accounting for the slenderness effect in columns is still valid except for the

calculation of flexural rigidity. The flexural rigidity should be calculated based on a modified modulus

of elasticity and the favourable effect of lower creep.

For beams and slabs, the use of the equivalent rectangular stress block (ACI 318 procedure) appears

to be adequate. Although it is not as conservative as for normal strength concrete, the compressive

Figure 3. Typical stressstrain relationships for steel bars



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strain limit of 0003 is still applicable. Likewise, the equation for minimum steel ratio as given in the1995 edition of the ACI building code already accounts for the increased tensile strength of higher

strength concrete and can be used without modification.

In the calculation of deflection, changes are needed in the expressions for modulus of elasticity. As

explained earlier, the code expression overestimates the modulus of elasticity for concrete strengths

exceeding 40MPa. The procedure for calculation of long term deflections also needs modification.This is to allow for the inclusion of the lower creep coefficient and diminished effectiveness of

compression reinforcement in high strength concrete beams.

1.5. Review of experimental work

Whereas several research projects have been concerned with the experimental behaviour of

reinforced concrete members with high strength concrete and high yield steel, very few have applied

loading and boundary conditions relevant to earthquake response. Furthermore, none of the previous

projects addressed the range of concrete compressive strength, longitudinal steel yield stress,

transverse steel yield stress and confining steel spacing comprehensively. It is, though, noted that such

early studies were most valuable in establishing trends and alerting designers to serious issues whichmay affect seismic safety. In Table II, the ranges covered by tests conducted by various researchers are

presented. The light shading indicates testing under axial force only (including eccentric loading),

whilst the darker shade indicates combined axial-flexural loads. Blank areas were not previously

tested.

Table III. Range of test parameters for confining steel

fy (MPa)

fc (MPa) 300500 500700 700900 9001100 b1100

6080 c,e,f,l f d,f a a,d,l,m80100 c,e,l,q o,p, k,q l

100120 b,c,d,e,g,j,q a,p d,j,q,r h d,nb 120 b,e,g,h h r h e,n

a. Ahmad and Shah9

b. Al-Hussaini et al.10

c. Bjerkeli et al.11

d. Nagashima et al.12

e. Muguruma et al.13

f. Cusson and Paultre14

g. Razvi and Saatcioglu7

h. Razvi and Saatcioglu15

i. Sun et al.16

j. Azizinamini and Kuska17

k. Kimura et al.18

l. Li et al.19

m. Tanaka et al.20

n. Sugano21

o. Bayrak and Sheikh22

p. Galeota and Giammatteo23

q. Muguruma and Watanabe24

r. Nishiyama et al.25

Table II. Range of test parameters for steel and concrete

fy (MPa)

fc (MPa) 300500 500700 700900 9001100 b1100

6080 a,c,d,f,l,m, d d m

80100 c,k,l,r k,p k 100120 b,c,d,j,q,r b,d,p d b120 b,r b



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With regard to confining steel characteristics and spacing tested, as identified from the published

literature, these are indicated in Table III.

It is clear from the above that there are considerable gaps in testing results of relevance to seismic

response, which requires attention prior to attempting to derive comprehensive design guidance. This

is further emphasized by the observation that some of the above tests were conducted under eccentric

axial load to represent combined axialflexural testing. Notwithstanding other objectives that the

researchers may have had in mind, this type of testing is considered by the authors to be not strictly

relevant for seismic assessment purposes. There are several behavioural considerations supporting the

latter statement, amongst which is that levels of axial load at maximum moment are so high that the

mobilized confining stresses are rather unrealistic and unrepresentative of the seismic response of

structures. A comprehensive testing programme on beamcolumn members was recently completed at

Imperial College (Elnashai et al.,26 Goodfellow27), with the range of test parameters summarized in

Table IV.

Table IV. Testing programme at Imperial College (Goodfellow27)

fc fyl fyt s (mm)

500 3550

500 785 5070

1300 5070

500 3550

685 785 5070

1300 5070

500 3550

70 100 130 785 785 5070

1300 5070

500 3550

900 785 5070

1300 5070

500 3550

1300 785 5070

1300 5070



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Tests on 92 members with concrete strength varying between 60 and 130MPa, steel yield strength

between 500 and 1300MPa with two stirrup spacing and two axial load levels under cyclic and

monotonic transverse loading have been completed. The results are still being processed. However,

one of the early interesting observations is that, in contrast to most of the observations in the literature,

hoops are rather lightly stressed and therefore there is no justification for the use of high yield steel as

confining reinforcement.

1.6. Seismic performance of high strength buildings

There are very few studies on high strength materials at the structure level. The study by Kateinas,28

summarized in Elnashai,29 was one of the first detailed analytical investigations into high strength

concrete buildings. It used the simplest approach possible for the analysis of a suite of buildings with

different concrete strength and steel yield. This comprised keeping the member size constant and

changing the material properties. Another feature of this work was the use of trilinear stressstrain

models for high strength concrete and multi-surface plasticity steel modelling. In the former case, the

descending branches were significantly steeper than the model employed in the current study.

Moreover, the higher the strength of concrete the steeper was the third segment of the model. Theabove three features of that study affected significantly the results obtained, hence the conclusions

drawn.

By keeping the section dimensions and ratio of reinforcement constant, many of the structures ended

up with response typical of strength design as opposed to ductility design, as well as sections being

over-reinforced. The consequence is to decrease the ductility of structures using high yield steel. This

was aggravated by the use of a non-ductile shape for the constitutive model representing the available

high yield steel. The feature of using severely dipping descending branches for concrete lead to the

structures behaving in a highly non-ductile fashion.

The approach used and models employed account comprehensively for the behaviour observed. It

was noted in that study that high strength structures exhibit very low levels of ductility and behaviour

factors nearing unity. In common though with the current study, the range of concrete and steel for

viable seismic design was identified. Therefore, the study by Kateinas,28

reported by Elnashai,29

represents the expected behaviour of high strength concrete structures if the confined concrete

behaviour is still non-ductile and the steel stress ratio (ultimate-to-yield) is near unity. Higher levels of

ductility would have been obtained though if the over-reinforced sections had been re-designed.

1.7. Scope of work

The main objective of this study is to investigate the seismic performance of high rise high strength

concrete buildings designed according to existing code provisions. To be able to make a realistic

comparison between structures designed using different grades of concrete, a number of equivalence

criteria are proposed here.

Ten 24-storey structures with the same overall dimension and geometry, sized according to the

criteria discussed in Section 2, were analysed elastically and designed according to the provisions of

the Uniform Building Code.30 Every effort was made to have realistically designed and detailed

structures. A comparative cost analysis of these structures was performed to determine their relative

cost effectiveness. Inelastic analyses were then performed on these structures using the program

ADAPTIC (Izzuddin and Elnashai31). Two sets of inelastic analysis were undertaken: first a static

pushover analysis of the structure was performed, followed by a set of dynamic analyses using three

artificial accelerograms scaled to the design ground acceleration (PGA) and twice the design PGA.



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2. EQUIVALENCE CRITERIA FOR DESIGN

2.1. Statement of the problem

A number of studies (Schmidt and Hoffman,32 Moreno and Zils,33 Smith and Rad,34 Webb,35) have

dealt with comparing the cost of normal strength concrete with high strength concrete based onequivalent strength. Most of these studies compared the cost of columns. For a given axial load,

different combinations of column sizes, concrete strength and steel ratio were obtained. The different

column sections were then compared based on the total in-place cost.

Comparing columns based on equivalent axial strength does not put any limit on the size of the

members. Since only the axial load is being considered, the most economical column is usually one

that has the smallest dimension possible and the minimum amount of steel reinforcement (Russell, 36).

In practice, such sections are only feasible in cases where the column moments are negligible and

where stiffness or drift limits are not of concern.

For columns used as part of a lateral load resisting system in high seismicity areas, the equivalent

strength criterion cannot be used. Drift limits and stiffness requirements will dictate the minimum size

of the members. Furthermore, there is no strict equivalence in strength between sections, if both axial

force and moments are considered together. Equivalence in terms of strength can only be defined ifaxial load capacity or moment capacity or taken separately. The shape of the interaction diagram for

axial load and moment capacity changes for different grades of concrete and steel. An example of this

is shown in Figure 4. The interaction diagrams illustrated are for three different columns, using

different grades of concrete, with the same axial load capacity. The difference in moment capacity is

noteworthy.

The constant member size equivalence criterion used by Kateinas28 and Elnashai29 bypass the

above problem, but poses others. On the one hand, which structure is more economical is a foregone

conclusion. Also, use of design expressions intended for normal concrete is not necessary. However,

the ensuing structures are of disparate strength and stiffness, hence they have to be analysed and

assessed using a constant supply-to-demand ratio, not constant input (demand).

Figure 4. Interaction diagram for columns with equal axial capacity



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As a means of obtaining a set of realistic equivalent structures using different grades of concrete,

the equivalent stiffness criterion is suggested herein.

2.2. Equivalent stiffness criterion

In the equivalent stiffness criterion, the stiffness and width-to-depth (b/h) ratio of the individual

members of the structure are maintained for different concrete strengths. Since the modulus of

elasticity of the concrete increases with increasing strength, the sizes of the sections have to be reduced

to maintain a constant stiffness as higher strength concrete is used. This will result in structures with

approximately equal overall stiffness.

The stiffness of the sections used in implementing the equivalent stiffness criterion (and also in the

analysis) is computed based on 50% of the gross section stiffness. This is in accordance with the

recommendation of SEAOC.37 It was suggested that this value is generally accepted to adequately

reflect the effective stiffness of a cracked member. This reduced section stiffness is applicable to all

members of the lateral load resisting system. According to SEAOC,37 section stiffness more than this

limit can be used if substantiated by a rational cracked-section analysis. If drift calculations are

critical, stiffness calculations based on cracked sections and transformed steel area can be employed.

Although this might seem to be more accurate than taking a percentage of the gross section stiffness, itcan be an excessive refinement considering the approximations involved in the mathematical model.

Many factors that might have a considerable effect on overall drift are usually not considered in an

elastic analysis. The relative degree of cracking of the different members of the structure, diaphragm

flexibility, foundation flexibility and joint stiffness are some of the factors that are not usually included

in the analytical model.

The equivalent section is computed by equating the EI of the section used for the normal strength

concrete with the one for the higher strength concrete, while maintaining a constant b/h ratio.

Rearranging the equation gives

b1 E0

E1

1a4b0 1

h1 E0

E1

1a4h0 2

where

b0, h0, E0 are the width, height and modulus of elasticity of the original section,

b1, h1, E1 are the width, height and modulus of elasticity of the equivalent section.

Preliminary analysis and design of the framewall structure (configuration described in Section 31below) has shown that most of the main lateral force resisting elements of the higher strength concrete

structures have minimum reinforcement. On the other hand, the structure designed with normal

strength concrete has many elements that are quite heavily reinforced. Indeed, strength governed thedesign of the normal strength structure. From practical design considerations, further reduction in the

size of sections in high strength concrete structures is still possible, as long as code drift requirements

are satisfied. This is explained further below.

2.3. Reduced stiffness criterion

The reduced stiffness criterion proposed here is employed for further reducing the size of the



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sections for high strength concrete structures. This criterion involves reducing the stiffness of the

members proportionally until the limiting drift of 002/Rw is reached. This entails an iterative processof reducing the size of the members, applying the appropriate loads and then analysing the structure

until the drift limit is reached. A drift of 002/Rw is the code limit beyond which P-delta effects aredeemed to be significant.

The underlying hypothesis of this criterion is that serviceability considerations rather than strengthwill govern the design of high strength high rise concrete structures. Indeed, it was noted by Ghosh and

Saatcioglu6 that it is quite common for a structural engineer to consider and specify high strength

concrete for its stiffness rather than for its strength.

3. ELASTIC ANALYSIS AND DESIGN OF MODEL STRUCTURES

3.1. Configuration of model structures

A model structure designed using normal strength concrete fHc = 35MPa) is the basis for allequivalent high strength concrete structures. Equivalence, in the sense of the criteria defined in Section

2, is the backbone of all the comparative analysis presented hereafter. The configuration of the 24-

storey frame-wall structure is shown in Figure 5.

The sizes of the beams are uniform throughout the height of the building, while the sizes of the

Figure 5. Elevation of 24-storey frame-wall model structure



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columns and shear wall change every eight storeys. The concrete strength and modulus of elasticity for

the different model structures is given in Table V.

The member sizes of the basic model structure fHc = 35MPa) and the corresponding equivalent highstrength concrete structures based on the equivalent stiffness criterion are given in Table VI. It should

be noted that the criterion was applied consistently to all the sections used in the building. The length

of the shear wall also changes because the criterion maintains a constant width-to-depth (b/h) ratio.

Indeed, if the length of the wall is held constant, the section that will give an equivalent stiffness will

have a width that is smaller than the practical minimum dimension of about 300mm.

Further reduction in member size is attained using the reduced stiffness criterion. Table VII shows

the member sizes for these structures. A slight deviation from the constant width-to-length ratio of the

shear wall is allowed here to be able to maintain a realistic minimum wall thickness of 300mm. The

reference code of the structures obtained using the reduced stiffness criterion is suffixed with an R.

Another scenario that is considered here is the possibility of using only one grade of steel throughout

a structure. This might be a consideration in cases where ease of construction is important. This

minimizes confusion due to the use of different steel grades for reinforcing bars of the same size on

site. To this end, a subset of structure, with the same dimension and reinforcing steel as the three model

structures with reduced stiffness, is included for inelastic analysis using one grade of steel (highest

Table V. Modulus of elasticity for different grades of concrete

Model No. N35 N80 N100 N120 N80R N100R N120R

fHc MPa) 35 80 100 120 80 100 120Ec (MPa) 27806 36590 40095 43264 36590 40095 43264

Table VI. Member sizes based on the equivalent stiffness criterion (mm)

Model No. N35 N80 N100 N120

Slabs 125 115 110 110Beams 400 800 375 750 365 725 350 700Columns 1st 8th 900 900 850 850 825 825 800 800Columns 9th 16th 800 800 750 750 725 725 700 700Columns 17th 24th 700 700 650 650 625 625 600 600Shear wall 1st 8th 450 8000 425 7500 400 7250 400 7000Shear wall 9th 16th 400 8000 375 7500 350 7250 350 7000Shear wall 17th 24th 350 8000 325 7500 300 7250 300 7000

Table VII. Member sizes based on the reduced stiffness criterion (mm)

Model No. N80 N100 N120

Slabs 115 110 110Beams 335 675 325 650 315 625Columns 1st 8th 750 750 725 725 700 700Columns 9th 16th 675 675 650 650 625 625Columns 17th 24th 600 600 575 575 550 550Shear wall 1st 8th 375 6700 365 6500 365 6400Shear wall 9th 16th 325 6700 325 6500 325 6400Shear wall 17th 24th 300 6700 300 6500 300 6400



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steel grade in the particular structure) throughout. The reference code of these structures is suffixed

with an RS.

3.2. High strength concrete and steel combination

The Uniform Building Code30 does not allow the use of steel with yield strength in excess of

415MPa in earthquake resisting elements. The reason given by the commentary is the problem with

bonding. However, the commentary suggests that the use of high strength concrete with high yield

steel can compensate for the reduction in available bond strength. In this study, the use of high yield

steel in combination with high strength concrete is imperative because of problems with congestion in

the smaller beam sections of the reduced stiffness structures. Based on the SEAOC commentary, 37 it

seems that the code limit on the yield strength of reinforcing steel is an interim provision and further

research will substantiate the use of high yield reinforcing bars with high strength concrete.

The basis for the choice of the steel grade for a given structure (concrete strength) is the steel ratio

(based on preliminary analysis) of the beam section. The reinforcing steel ratio was limited to 0025.This is the code limit and is intended to ensure adequate rotational capacity before concrete crushing.

A summary of the different model structures with the grade of concrete and steel used is shown in

Table VIII.

3.3. Design criteria

The model structures are designed and detailed according to the pertinent provisions of the Uniform

Building Code.30 As required by the code, to provide additional safeguard against total collapse, the

frame is designed to resist at least 25% of the base shear without the benefit of the shear wall. It is

assumed that wind load will not govern.

The following load combinations are considered:

U 14DL 17LL 3

U 14DL LL EQ 4

U 09DL EQ 5

The program SAP9038 was used to perform the 2D elastic analysis of the different structures. SAP90

Table VIII. Summary of structures considered

fy (MPa)

Model f'c (MPa) Beams Columns Shear Wall Hoops Remarks

N35 35 415 415 415 415 Basic model structure

N80 80 415 415 415 415 Equivalent stiffnessN100 100 415 415 415 415 Equivalent stiffnessN120 120 415 415 415 415 Equivalent stiffnessN80R 80 600 415 415 415 Reduced stiffnessN100R 100 700 415 415 415 Reduced stiffnessN120R 120 800 415 415 415 Reduced stiffnessN80RS 80 600 600 600 600 One steel grade onlyN100RS 100 700 700 700 700 One steel grade onlyN120RS 120 800 800 800 800 One steel grade only



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is a general structural analysis program evolved from the SAP series of programs developed at the

University of California, Berkeley. Based on the results of the analysis, detailed design was performed

using spreadsheet programming.

3.4. Summary of design results

The schedule of reinforcement and detailed drawings of the main lateral resisting elements of the ten

model structures is given in Laogan and Elnashai.39 Some noteworthy observations regarding the

results of the elastic analysis and design are as follows.

. The total base shear to which the dynamic lateral forces is scaled to about 3% of the total weight.

This is the code minimum base shear.

. P-delta effects need not be considered because the drift is below the 002/Rw limit set by the code.

. The choice of the grade of steel with the different grades of high strength concrete was validated

by the results of the design. Most of the beams are within the steel ratio limit set by the code.

. The strong-column weak-beam provision of the code did not govern the design of the columns

because all the columns are part of exterior joints, hence the capacity of two columns combined is

compared with a single beam.. Consistent with the preliminary analysis, the columns and shear wall of the high strength concrete

model structures have minimal reinforcement. The increase in concrete strength has more than

compensated for the decrease in section size.

. The shear stress level in the columns and shear wall is relatively low. The design of the sections at

the lower levels was governed by axial load.

. In the design of the shear wall for model structure N35 (normal strength concrete), the amount of

steel required was governed by the requirement to limit the axial stress. This resulted in heavy

steel reinforcement in the shear wall of the lower half of the structure.

. As the strength of concrete increases, the amount of confinement steel required in the shear wall

boundary elements increases substantially. The UBC equation for confinement is a function of the

size of the boundary element, the strength of concrete and the yield strength of steel. In many

instances, the number of vertical bars in the boundary element was governed by the need to usemore confining ties.

4. COMPARATIVE COST ASSESSMENT

4.1. Economics of high strength concrete

High strength concrete can support an axial compressive load at lower cost than normal strength

concrete. The use of smaller size columns has significant advantages. In the lower storeys of high rise

buildings, the area occupied by the columns is usually quite substantial. It was noted in the Chicago

task force report referred to in the ACI Committee 363 Report2 that the potential number of stories of a

high rise building is limited by the required large columns. The use of smaller high strength concrete

columns would provide more flexibility for the architectural layout and at the same time increase the

area of usable floor space.

Designers of high rise buildings have recognized early on the economic advantage of using higher

strength concrete in the columns. Apparently, to maximize the potential cost savings, there is a

tendency to use lower grade of concrete or other materials for the floor system. This practise has

prompted the ACI Committee 318 to include a provision in the building code requiring special

precaution in cases where the column concrete strength exceeds floor system concrete strength by

more than 40%. This provision is based on a study by Bianchini et al.40 of the effects of having a



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different grade of concrete in the floor system on the column strength. For buildings built in high

seismic hazard regions, this practice of using different grades of material for the columns and the floor

system can be a source of potential problems. In addition to the likelihood of confusion during

construction, the effect of having different materials around the beamcolumn joint might have

adverse consequences on its seismic performance. Further research on this should be undertaken to

confirm the acceptability of this practice.From the point of view of the cost of raw materials, the use of high strength concrete in beams and

slabs might not be as effective as for columns. The primary advantage of having high strength concrete

beams is the savings in dead load due to reduced section dimension.21 However, Schmidt and

Hoffman32 pointed out that the additional reinforcement required might offset the savings achieved

from reducing the section sizes. However, the indirect benefits of having a shallower beam section can

be quite significant. For a specified headroom, a shallower beam results in a lower storey height. This

translates into savings in the amount of concrete due to reduction in the overall height of the building.

In addition, the amount of exterior cladding required, which is usually a large portion of the total cost

of a building, is reduced. Lower storey height can also yield savings from the reduced area of finishes

required and the shorter runs of utility lines and HVAC ducts.

4.2. Material estimate

The cost comparison involves only the three main materials that are needed for reinforced concrete

construction; concrete, reinforcing steel and formwork. Detailed material estimates were undertaken

based on the design of the ten model structures as presented in Section 3. The concrete volume,

indicated in the material estimate, includes all the concrete needed for the structural members of the

frame-wall and the ten metre slabs attributed to the frame. The reinforcing bars estimate covers all the

main longitudinal reinforcement (both horizontal and vertical), ties hoops, and stirrups. The required

bends and hooks for anchorage were also included. The formwork area covers the three faces of the

beams, all four faces of the columns and shear walls up to the bottom of the beam, and the underside of

the slab between the beams and columns. The material estimates for the different model structures are

summarized in Table IX.

It is noted that the volume of concrete and the area of formwork decrease with increasing concretestrengths. This is expected with both the equivalent stiffness and reduced stiffness criterion. In both

criteria, the member sizes decrease with increasing concrete strengths. In terms of the amount of steel

reinforcement, there is a substantial differences (about 25%) between the normal strength concrete

structure (N35) compared with the other high strength concrete structures (Figure 6). Further

examination of the data showed that this difference arises from the vertical reinforcement in the

columns and shear walls. This can be attributed to the fact that in structure N35 the high axial loads in

the lower storeys have to be resisted by a combination of the concrete and a large amount of steel

reinforcement. On the other hand, in the high strength concrete structures, the high axial loads were

resisted primarily by the concrete and the minimal steel reinforcement. This emphasizes the advantage

of using high strength concrete in axial load-carrying elements.

It is also observed from Figure 6 that the amount of steel used as hoops and ties increases slightly

with increasing concrete strength. Nevertheless, the reduction in the longitudinal reinforcement more

than compensates for the increase in the hoop reinforcement.

4.3. Material prices

While every effort has been made to acquire a representative unit cost of the individual materials,

regional practice significantly affects prices and availability. Unit cost for the high strength concrete is



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especially difficult to determine since the concrete suppliers surveyed were reluctant to give price

information without prior knowledge of project requirements.

The price of the materials used in this cost analysis were based on sources from different

geographical locations. This limitation should be taken into consideration when interpreting the

results. The unit cost of the high strength concrete were taken from the literature4 and updated using

the US Bureau of Labour Statistics producers price index for concrete. The cost of high yield steel

was based on the prevailing prices of commercially available grades in Japan (M. Iwata, private

communication). For the formwork and normal grade of steel and concrete, unit costs were based on

the prices in the CESMM3 Price Database41 and adjusted for inflation. However, since the volumes of

the separate ingredients are given above, the total cost comparison can be readily updated.

The unit cost of the different materials is tabulated in Table X. All the prices are in-place unit costs.

Table IX. Summary of material estimates

Model Materials Beams Columns Shear Wall Slabs Total

Concrete (m3) 124 99 246 90 559N35 Formwork (m2) 680 369 1 227 717 2 993

Gr.415 Rebar (kg) 28 288 24 173 47 299 4 925 104 686Concrete (m3) 113 87 216 83 499

N80 Formwork (m2) 661 353 1 155 719 2 888Gr.415 Rebar (kg) 28 367 25 337 22 288 4 545 80 537


Gr.415 Rebar (kg) 29 184 27 183 20 784 4 355 81 507


Gr.415 Rebar (kg) 30 501 25 861 21 087 4 361 81 809

Concrete (m3) 96 71 172 83 421N80R Formwork (m2) 615 327 1 033 723 2 698

Gr.415 Rebar (kg) 4 917 19 783 19 014 4 568 48 281Gr.600 Rebar (kg) 21 259 21 259

Concrete (m3) 90 65 165 80 400N100R Formwork (m2) 602 318 1 004 724 2 648


Concrete (m3) 85 61 162 80 387N120R Formwork (m2) 580 309 990 724 2 604


Concrete (m3) 96 71 172 83 421N80RS Formwork (m2) 615 327 1 033 723 2 698

Gr.600 Rebar (kg) 26 176 19 783 19 014 4 568 69 540

Concrete (m3) 90 65 165 80 400N100RS Formwork (m2) 602 318 1 004 724 2 648

Gr.700 Rebar (kg) 26 129 20 333 19 226 4 374 70 061

Concrete (m3) 85 61 162 80 387N120RS Formwork (m2) 580 309 990 724 2 604

Gr.800 Rebar (kg) 26 252 20 235 19 944 4 378 70 808



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The cost of labour and plant/equipment time are included in the prices. For the reinforcing bars and

formwork, an extra component that accounts for material wastage is included in the price. The rental

cost for the shoring equipment is included in the cost of the formwork.

It is noteworthy that the price of high yield steel is double that of grade 415 steel, while the cost of

the 80Mpa high strength concrete is only about 18% more than 35MPa concrete.

4.4. Results of cost comparison

The total cost of the different model structures, based on the unit costs shown on Table X, are

summarized in Table XI.

The most cost-effective structure appears to be N80R. This is attributable to its using only a minimal

amount of the expensive high yield steel (used in the beams only) while at the same time taking

advantage of the large axial load carrying capacity of the high strength concrete to reduce the vertical

steel reinforcement. Figure 7 shows a bar chart of the cost broken down into the different components.

As a group, the structures designed using the equivalent stiffness criterion (N80N120) show an

increase in the cost of the concrete component as the strength of the material increases. The cost of the

Figure 6. Steel reinforcement for the different structures

Table X. Unit cost of materials

Materials Unit cost

35MPa Concrete (US $/m3) 1400080MPa Concrete (US $/m3) 16500100MPa Concrete (US $/m3) 19600120MPa Concrete (US $/m3) 24800Grade 415 Steel (US$/kg) 096Grade 600 Steel (US$/kg) 186Grade 700 Steel (US$/kg) 193Grade 800 Steel (US$/kg) 200

Formwork (US $/m

2

) 3400



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formwork remains relatively stable. The cost of the steel reinforcement is almost the same for all the

high strength concrete structures and is substantially less than that for the normal strength concrete

structure. The group designed using the reduced stiffness criterion seems to be the most economical.

The cost of concrete components increases with increasing concrete strength but is generally less than

that for normal strength concrete structures. The cost of formwork, as in the previous group, remains

approximately the same, while the cost of the steel component shows a slight increase with increasing

Table XI. Total cost of the structures and percentage savings relative to N35

Cost in US$

RebarsSavings*

Model Concrete Formwork Gr.415 Gr.600 Gr.700 Gr.800 Total (%)

N35 78 279 101 762 100 499 0 0 0 280 540 000%N80 82 369 98 195 77 316 0 0 0 257 880 808%N100 90 949 96 345 78 246 0 0 0 265 541 535%N120 110 493 94 282 78 537 0 0 0 283 312 099%N80R 69 441 91 734 46 350 33 404 0 0 240 930 1412%N100R 78 460 90 039 46 756 0 41 220 0 256 474 858%N120R 96 071 88 522 47 180 0 0 43 325 275 098 194%N80RS 69 441 91 734 0 123 207 0 0 284 383 137%N100RS 78 460 90 039 0 0 135 218 0 303 717 826%N120RS 96 071 88 522 0 0 0 141 617 326 209 1628%

* Savings are relative to the cost of N35 (-ve means higher cost).

Figure 7. Cost of the different components of the structures



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concrete strength. Structures using only one grade of steel appear to be the least economically viable.

This is not surprising considering the large difference in cost between normal and high yield steel. As

steel technology improves and demand increases, reduction in the price of high yield steel is expected.

Table XII shows the relative savings that can be attained with the use of high strength concrete, if the

cost of high yield steel is reduced by 30%. These figures contrast with those shown in Table XI.

At this cost level, it is noticeable that even the structures using only one grade of steel become

economically feasible. Moreover, the use of high strength concrete at this price level can result in

substantial savings. This suggests that the cost effectiveness of a high strength structure depends to a

large extent on the cost component attributable to the steel reinforcement.

Contrary to conclusions of Schmidt and Hoffman32 for high strength concrete columns, it appears

that the optimum concrete strength is not necessarily the highest strength available. The concrete

strength that will give the most cost-effective structure is the one that can allow for the largest

reduction in the cost of the steel component, while having the least increase in the cost of the concrete

component. The optimum concrete strength depends on the relative cost of the different grades of

concrete and steel.

The results of this cost analysis also serve to show that substantial cost savings can still be attained,

even with the use of high strength concrete in the beams and slabs where they are less effective. The

key to having an economical high strength concrete structure is the proper choice of the concrete and

steel combination to use.

5. INELASTIC ANALYSIS: PROCEDURES AND RESULTS

5.1. Program, models and methods

The main analytical tool used for this study is the program ADAPTIC.31 ADAPTIC is an adaptive

static and dynamic analysis program developed for the nonlinear analysis of steel, concrete and

composite frames. The program is capable of predicting the large displacement static and dynamic

behaviour of elastic and inelastic plane and space frames. The concrete model developed by Martinez-

Rueda and Elnashai42 was employed for the analysis. This model was developed for normal strength

concrete, hence the shape of the descending branch for unconfined concrete is distinct from that

observed for high strength concrete. However, this discrepancy will only affect the cover response,

which is not a critical aspect, especially on a whole structure level. Moreover, the currently observed

dispersion in the results from stressstrain relationships developed specifically for high strength

concrete lends weight to adopting well-verified normal concrete models, at least for the time-being.

The bilinear steel model with kinematic strain hardening was adopted for modelling the reinforcing

steel. This follows earlier studies by Kateinas28 and Elnashai and Izzuddin,43 where it was indicated

that the difference between bilinear representation and more refined models is not large.

For both the static and dynamic analysis, a gravity load equivalent to 1 4 times the combined deadand live load was applied as an initial load. The input motions used for the dynamic analysis are three

of the EC8 series of artificial accelerograms.44 These records are consistent with both the Eurocode 8

and UBC elastic response spectrum. The plot of the acceleration time-history and the response

spectrum of one of the records is shown in Figure 8. It is noted that the artificial records were generated

Table XII. Percentage savings relative to N35 with reduced HYS prices

Model N35 N80 N100 N120 N80R N100R N120R N80RS N100RS N120RS

Savings (%) 000 808 535 099 1770 1299 657 1185 622 114



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for soil class B (firm soil) of the Eurocode 8, which is distinct to the soil class assumed in the design.

The difference is higher amplification in the long period range, hence the records are conservative.

5.2. Static pushover analysis

The static pushover analysis involves subjecting a structure to a proportionally increasing lateral

load, until the ultimate limit state of the structure is reached, while controlling the top displacement. In

this study, the response of the structures at the yield and ultimate limit state is investigated. There are a

number of proposals for the definition of the yield limit state (YLS). The yield point of the structure

defined based on an equivalent elasto-plastic system with reduced stiffness, evaluated as a secant

through 75% of the maximum, is adapted for this study. This definition is graphically illustrated in

Figure 9. On the other hand, the ultimate limit state (ULS) is based on a global criterion of the

attainment of 3% interstorey drift. At this level of drift, it is generally accepted that structures wouldhave suffered major structural and non-structural damage. Further refinement of the limit states is not

Figure 8. EC8-1 acceleration time history and response spectrum

Figure 9. Yield point based on equivalent elasto-plastic system



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necessary since the primary objective is to compare the different structures. Table XIII summarizes the

yield and ultimate displacement for all buildings.

The displacement defining the ULS averages about 2100mm. This is equivalent to 27% of the totalheight of the building. Figure 10 shows the forcedisplacement curve for the normal strength and the

equivalent stiffness structures. It is noticeable that the initial stiffness of the structures is the same. This

feature is also exhibited in the reduced stiffness structure, thus validating the equivalence criteria used.

This also confirms that imposing equivalent stiffnesses at the member level will give a structure with

the same overall stiffness.

Plastic hinging, as referred to in this study, is defined as having a strain at the outermost layer of

steel reinforcement exceeding the yield strain of the material. A typical hinging pattern of an HSCstructure at the ULS is shown in Figure 11. Analysis of the plastic hinge formation under a static load

has shown that inelasticity develops mostly in the beams of the structures. This is in good agreement

with the code principle of having energy dissipation concentrated in the beam. For this type of loading,

the columns appear to be well protected from hinging. This is consistent with the code strong-column

weak-beam principle.

The level of overstrength is defined here as the ratio of the capacity of the structure, based on the

static pushover analysis, to the code-defined design base shear. The calculated overstrength for the

different structures is given in Table XIV.

All of the structures, except N35, have overstrengths in the range of 4 to 5. The higher level of

overstrength in the normal strength structure is attributable to the large amounts of steel reinforcement

in the lower storey shear walls, which were provided to resist the axial force. This steel provides

considerable flexural resistance under transverse load.

Table XIII. Yield and ultimate displacements (mm)


Dy 644 398 405 417 636 768 896 649 791 938Du 1880 2100 2100 2120 2140 2080 2120 2120 2060 2120

Figure 10. Forcedisplacement curve for N35, N80, N100 and N120



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5.3. Inelastic dynamic analysis

For each of the ten structures, dynamic analysis was performed with three artificial records scaled tothe design ground acceleration (PGA) and twice the design PGA, corresponding to 04 and 08g,respectively. As a measure of global response characteristics, the displacement and total base shear

time-history and the interstorey drift ratio from the analyses are examined. For the purposes of

discussion, N80, N100 and N120 constitute group E (based on the equivalent stiffness criterion),

N80R, N100R and N120R group R (based on the reduced stiffness criterion) and N80RS, N100RS and

N120RS group RS (based on the reduced stiffness criterion with one steel grade only).

It was observed that for buildings in the same group and subjected to the same input motion, the

shape of the displacement and base shear time-history was similar. This suggests that structures in the

same group responded in a fairly similar manner, confirming the effectiveness of the criteria developed

for generating the structures. To illustrate the point, the top displacement and base shear time-history

of the R structures is shown in Figure 12. A summary of the maximum top displacement of the

different structures is given in Table XV.

The maximum top displacement of the structures within the same group shows a slight increase as

the concrete strength increases. This is consistent with the loaddisplacement curves generated under

static loading. The higher strength concrete experiences slightly more softening beyond a certain strain

in the initial ascending branch of the curve. This is partially attributable to the cracking of the concrete

cover. The size of the cover is constant in all the structures, but the sizes of the members decrease with

increasing concrete strength. The ratio of the cover to the total depth of the structural member is

Figure 11. Typical plastic hinge formation of HSC at ULS under static loading

Table XIV. Structure overstrengths


Over-strength 629 451 447 442 420 467 493 425 477 510



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Figure 12. Displacement and base shear time history of R-structures subjected to EC8-1

Copyright

1999Joh

nWile

y&Sons,Ltd.

Struct.DesignTallBuild.8,171204(1999)

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therefore larger in the higher strength concrete structures. In all instances, the maximum top

displacement of structures in the R and RS groups is greater than that for structures of the sameconcrete strengths in group E. This is expected because of the higher overall stiffness of the structures

in group E. The top displacement of the normal strength concrete structure N35 is comparable with the

maximum values for the structures in group E.

The interstorey drift ratio is the basis for the global failure criterion defined earlier. A sample of the

plot of distribution of maximum drift ratio along the height of the structure is shown in Figure XIII.

It is observed that the maximum drift ratio generally increases with height. It is important to note

that the maximum drift ratio at all floor levels does not necessarily occur at the same instant; as such,

Table XV. Maximum top displacement (mm)

EC8-1 EC8-2 EC8-3

Model 04g 08g 04g 08g 04g 08g

N35 601 1074 618 1208 552 1103N80 610 1149 672 1334 600 1166N100 632 1161 687 1351 643 1220N120 646 1145 707 1342 668 1280N80R 779 1270 839 1157 787 1212N100R 803 1261 896 1228 813 1428N120R 847 1267 963 1393 857 1547N80RS 783 1262 839 1164 788 1220N100RS 813 1255 894 1250 819 1433N120RS 861 1328 950 1365 859 1557

Figure 13. Distribution of maximum drift ratio along a typical HSC structure



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the graph shown is not a profile of the displaced shape of the structure. The largest absolute value of the

maximum drift ratio that occurred along the height of the different structures is given in Table XVI.

It is noted from Table XVI that the use of different input motion yields a slightly different maximumstorey drift ratio. At a PGA of 08g, all the structures are still within the limiting drift ratio of 3%. It isobserved that although N35 has a higher level of overstrength, based on static pushover analysis, its

maximum drift ratio similar to that of the structures in group E is already near the 3% limit. This

indicates that higher mode response has a significant effect.

The plastic hinges formed in the different structures show that the number of column hinges

generally increases with increasing concrete strength. This is attributable to the higher strain levels in

the steel due to the increase in strength and modulus of elasticity of the concrete used. Although the

strong-column weak-beam provision of the code was satisfied in the design, the 120MPa concrete

Table XVI. Maximum interstorey drift ratio

ModelEC8-1 EC8-2 EC8-3

04g 08g 04g 08g 04g 08g


Figure 14. Plastic hinge formation of R structures subjected to EC8-10 8g



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structures in group R show substantially more hinging in the columns (Figure 14). The use of high

yield steel in the beams and normal grade steel in the columns of the R structures has an adverse effect

in terms of an increased number of column hinges. The structures in group RS, which used high yield

steel in the columns, had a few column hinges even at high levels of loading. On the other hand, the

spread of inelasticity in the shear wall is unaffected by the change in concrete strength. The use of

normal grade steel with high strength concrete does not cause any undesirable consequences in the

wall. For the beams, at twice the design PGA, yielding has occurred in almost all cases. This is

expected and desirable at this level of loading.

Difference in the characteristics of the input motion did not have an evident effect on the location of

plastic hinges in the structures. However, this does not give a conclusive indication that the local

response was not affected by the characteristics of the record. To verify probable local effects, the

curvature ductility demand of a number of beams at different locations in the structure was calculated.

There are various definitions of yield curvature, such as that of Paulay and Priestley,45 the yield

curvature to be 133 times the curvature at first yield. In the absence of a more rigorous analysis, thisdefinition was adopted for this study. The maximum curvature ductility demand in the beams of the

structure divided according to the criterion used in generating the structure are shown in Table XVII.

It is interesting to note that the maximum curvature ductility demand occurred in the normal

strength concrete structure and the equivalent stiffness HSC structures. The ductility demand at a PGA

of 08g is already on the high side, but is within the achievable ductility capacity of properly designedand detailed members. For the R and RS structures, the ductility demand is significantly less compared

with those of group E. This is partially due to the definition of the yield curvature. Since the R and RS

group structures use high yield steel in the beams, the strain that defined yield is much higher

compared with normal strength steel, hence a higher yield curvature and consequently a lower ductility

demand was observed.

Examination of the calculated ductility demand in the beams did not show any particular trend or

pattern, except for the consistently large difference in values mentioned above between E structures

and R and RS structures. The maximum values given in Table XVII did not occur at the same point nor

at the same instant. However, it is noticeable that many of the larger values occurred in the upper third

portion of the structure (higher mode effects). The variation in ductility demand, using different input

motion, is large. This indicates that in terms of local response the choice of input motion does have a

significant effect.

5.4. Behaviour factor

Behaviour factor (also termed response modification factor) is an important parameter in force-

based seismic design. It may be viewed as a global measure of force reduction due to inelastic effects.

Table XVII. Maximum curvature ductility demand in beams

Group E Group R Group RS

Model 04g 08g 04g 08g 04g 08g

EC8-1 311 715 168 399 171 387EC8-2 373 703 163 383 168 390EC8-3 233 609 165 305 167 312



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It may be estimated analytically from the equation (e.g. Elnashai and Broderick,46 amongst others)

q ac

ay6

where

ac is the peak ground acceleration at collapse

ay is the peak ground acceleration at yield

For each structure, determining the intensity of the ground motion at yield and collapse involves a

number of inelastic analyses. This requires a considerable amount of computational time and effort.

Based on the analysis with the records scaled to 0 8g, it is noticeable that at this acceleration level mostof the structures are already near the global ultimate limit state. Subsequent analysis has shown that

calculating the collapse acceleration using direct proportionality gives reasonably close valuescompared with those obtained from repetitive analyses. On the other hand, the yield acceleration can

be defined using the design loads with due consideration to the available overstrength as follows:

ay ad

Rw

OS

147

where

ad is the design peak ground acceleration

Rw is the behaviour factor used in the design

OS is the calculated overstrength based on static pushover analysis

The coefficient 14 in the denominator accounts for the load factor on the design base shear in UBC.The calculated overstrength shown in Table XIV is based on the unfactored design base shear. The

values of the calculated peak ground acceleration at yield and at collapse using the three different input

motion is given in Table XVIII.

In assessing the results given in Table XVIII, it should be noted that behaviour factors determined

analytically are highly dependent on the definition of the yield and ultimate criteria and the choice of

the input motion used in the analysis. For the purposes of this study, these calculated values are used as

Table XVIII. Calculated collapse PGA, yield PGA and behaviour factor

ac (g) ay (g) q

Model EC8-1 EC8-2 EC8-3 EC8-1 EC8-2 EC8-3 EC8-1 EC8-2 EC8-3

N35 0841 0950 0902 0150 0150 0150 561 633 601N80 0870 0982 0871 0107 0107 0107 813 918 814N100 0868 0954 0847 0106 0106 0106 819 900 799N120 0893 0970 0869 0105 0105 0105 850 924 828N80R 0934 0864 0980 0100 0100 0100 934 864 980N100R 0958 0911 0898 0111 0111 0111 863 821 809N120R 0984 0939 0876 0117 0117 0117 841 803 749N80RS 0963 0901 0980 0101 0101 0101 953 892 970N100RS 0959 0918 0869 0114 0114 0114 841 805 762N120RS 0997 0958 0853 0121 0121 0121 824 792 705



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a comparative parameter to give some insight into the relative energy dissipation and redistribution

capacity of the different structures.

Of all the structures considered, the normal strength structure has the lowest behaviour factor. This

is attributable to its having a much larger level of overstrength compared with the other structures. The

extra steel reinforcement provided for gravity loading in the normal strength concrete has apparently

affected its energy dissipation capacity. Its yield acceleration is much greater than that of the otherstructures, while the collapse acceleration is similar to that of others. The degree of overstrength in a

structure affects the yield acceleration more than the collapse acceleration. The behaviour factor for

the structures in group E is almost constant. On the other hand, the behaviour factor generally

decreases with increasing concrete strength for both R and RS structures (with very few exceptions).

Comparison of the calculated behaviour factors suggest that in terms of structural ability to respond in

the inelastic range, 80MPa concrete is the most effective for the given structural configuration.

5.5. Fourier amplitude spectra of response acceleration

The inelastic period of the structure is examined using the discrete Fourier amplitude spectra of the

response acceleration time-history at the top of the structure. The plot of the Fourier amplitude spectra

for structure N35 is shown in Figure 15. The first peak in Figure 15(a) at 0 51seconds and the secondpeak at 232seconds correspond to the second mode and the fundamental mode periods of thestructure, respectively. During the preliminary analysis, the elastic periods of the structures were

verified using an eigenvalue analysis. The inelastic period does not coincide exactly with the values

obtained from the eigensolution because of the softening effect due to cracking of the concrete.

However, there was a good one-to-one correspondence between the peaks observed in the Fourier

amplitude spectra and the periods obtained using the eigenvalue analysis.

As observed in Figure 15, the inelastic periods shift slightly to the right as the load is increased. The

shift in the fundamental mode period is much more than that in the second mode period. This is

because the displacement amplitude of the fundamental mode response is larger than that of the second

mode. The same trend is exhibited in all the other structures. It is also noted that the Fourier amplitude

spectra at a PGA of 08g become quite erratic. This is due to the high levels of inelasticity present at

this intensity of loading, where stiffnesses are constantly changing. For the case of a PGA of 08g,judgement has to be made on which peak in the Fourier amplitude spectra corresponds to the different

Figure 15. Fourier amplitude spectra for N35 with EC8-1: (a) 04g; (b) 08g



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modes of vibration. The inelastic periods corresponding to the fundamental and second mode of

vibration, for all the different structures, are indicated in Table XIX.

The inelastic periods of the structures within the same group are almost the same. As expected, the

structures in the R and RS groups have slightly longer periods. Variation of the inelastic period due to

the difference in input motion are as much as 30%. To compare the inelastic with the design periods,

the first and second mode periods obtained from the SAP90 analysis are given in Table XX.It should be noted that in the SAP90 analysis (eigenvalue analysis), the stiffness of the members was

calculated based on 50% of the gross section stiffness. Interestingly, the periods are in good agreement

with the inelastic periods at a design PGA of 04g. This partially validates the code recommendation ofusing a percentage of the gross section stiffness in determining the design period of the structure. It

Table XIX. Inelastic period (s) at PGA of 04 and 08g

04g 08g

Model Mode EC8-1 EC8-2 EC8-3 EC8-1 EC8-2 EC8-3

N35 1st Mode 237 220 212 288 290 3702nd Mode 061 067 056 083 070 068

N80 1st Mode 268 310 215 320 335 4302nd Mode 061 067 056 083 070 066

N100 1st Mode 279 280 200 380 320 3652nd Mode 062 068 056 085 070 066

N120 1st Mode 254 304 215 400 328 4202nd Mode 062 068 056 088 070 066

N80R 1st Mode 340 300 350 440 342 3922nd Mode 085 070 066 088 098 071

N100R 1st Mode 342 305 346 450 330 3642nd Mode 083 071 067 088 098 072

N120R 1st Mode 351 320 320 480 352 3702nd Mode 083 071 067 089 098 073

N80RS 1st Mode 322 318 392 400 345 480

2nd Mode 085 071 066 089 098 073N100RS 1st Mode 350 310 338 420 336 380

2nd Mode 086 071 067 089 098 072N120RS 1st Mode 372 318 340 415 380 376

2nd Mode 086 071 067 090 098 077

Table XX. Design periods obtained from SAP90

Model Period (s) 1st mode Period (s) 2nd mode

N35 287 073N80 277 070

N100 281 071N120 288 073N80R 344 086N100R 348 086N120R 352 086N80RS 344 086N100RS 348 086N120RS 352 086



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appears that it is possible to determine the inelastic periods of the structure by means of a much simpler

eigenvalue analysis, using appropriate member section stiffness. Further parametric studies should be

performed to correlate the percentage of the gross section stiffness to use, design acceleration andinelastic periods for different structural systems.

The relative contribution of the different modes of response depends on both the characteristics of

the structure and the input motion. To examine the relative contribution of the first and second mode

response, the amplitude of the Fourier spectra at these periods is compared. The ratio of the spectral

amplitude at the second mode period to the amplitude at the fundamental mode period is given in Table

XXI.

There is no noticeable trend in the ratio of the spectral amplitude of the second mode to the first

mode response. The most evident observation is that the second mode response predominates in almost

all instances. However, the fundamental mode response should not be discounted considering the

relatively significant contribution.

The results of this part of the analysis have some implications on displacement-based design. It was

indicated in Bommer and Elnashai47 that based on the current catalogue of earthquake records, reliable

displacement spectra can be derived only up to a period of three seconds. This apparently might cause

some problems for applications in high rise structures, whose fundamental period might exceed three

seconds. However, based on the results of this study, it appears that the predominant mode of response

of the high rise frame-wall structure considered is the second mode. If the second mode period is taken

as the effective period for displacement-based design, then there is no complication with the limit in

the period range of the displacement spectra. On the other hand, if the contribution of the fundamental

mode, which is relatively significant, is to be considered together with the second mode period in

deriving an effective period, then a procedure of relative weighting has to be developed.

6. CONCLUSIONS

In this paper, the seismic performance and cost effectiveness of high rise high strength RC buildings is

investigated. Based on the analyses performed, the following conclusions are supported by the results.

. The cost effectiveness of a structure depends to a large extent on the cost component attributable

to the steel reinforcement. Those structures using a large amount of expensive high yield steel

prove to be the least competitive in terms of cost. With the current prices of high yield steel, its use

should be limited to cases where the use of normal grade steel is not possible. However, a price

Table XXI. Ratio of the second mode to the first mode spectral amplitude

EC8-1 EC8-2 EC8-3

Model 04g 08g 04g 08g 04g 08g




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reduction in high yield steel of about 30% from current levels will make even structures that uses

only one grade of steel economically viable.

. Contrary to some of the studies performed on columns subjected to pure axial load, the optimum

concrete strength that will give the most economical structure is not necessarily the highest

concrete strength available. The concrete strength that can reduce the cost of the steel component,

while at the same time limit the cost of the concrete component, will result in the most costeffective structure. The optimum concrete strength depends on the relative prices of the different

grades of concrete and steel.

. In light of the common practice of using high strength concrete only in the columns of a structure,

it was shown that the use of high strength concrete in the beams and slabs, where they are less

effective, can still furnish substantial cost savings. Excepting any detrimental effect on seismic

performance that might be observed in the future, using just one grade of concrete in both the

columns and beams of a structure has the added advantage of eliminating the additional

construction procedure required by the code.

. Under static loading, high strength concrete structures have stable loaddisplacement curves. The

shape of the curve is similar to that of normal strength concrete structures. For this type of loading,

inelasticity developed mostly in the beams. The columns appear to be well protected from hinging

by capacity design regulations.. The level of overstrength in high strength concrete structures, calculated based on static pushover

analysis, is less than that of a normal strength structure. In a normal strength structure, the

additional steel reinforcement required to resist high axial loads provides extra lateral load

capacity, thus increasing overstrength.

. At the global level, based on the three response parameters examined (top displacement, base

shear and maximum interstorey drift ratio), the performance of high strength concrete structures

compares favourably with that of the equivalent normal strength concrete structure. There were no

indications that a properly designed high strength structure would behave any differently from the

equivalent normal strength structure.

. The choice of the grade of steel to use with high strength concrete is very important. Results from

dynamic analysis indicate that using normal grade steel for concrete strengths of up to 80MPa is

adequate. Beyond 80MPa, the use of normal grade steel with high strength concrete resulted insignificantly more hinging in the columns. The use of high yield steel in beams and normal grade

steel in columns should also be avoided.

. The maximum curvature ductility demand at the design and twice the design earthquake is 373and 715, respectively. These are well within the achievable ductility capacity of seismicallydesigned and detailed members. Moreover, the use of high yield steel with high strength concrete

significantly reduced the ductility demand on the members to about 1 63 and 383.. The use of different input motion, generated and scaled to fit the code design spectra, does not

have a significant effect on the global response parameter studied. The differences are well within

the limits of tolerances acceptable for seismic design and analysis. However, in terms of local

response, as in the case of the curvature ductility demand on the structural members, caution

should be exercised in assessing the results for design or analytical purposes. These parameters

appear to be highly dependent on the input motion used for the analysis.

. The calculated behaviour factors suggest that 80MPa concrete is the optimum concrete strength

for the given structural configuration, in terms of energy dissipation capacity. In general, higher

strength concrete structures have a larger behaviour factor compared with a normal strength

structure.

. For both normal and high strength concrete structures, the inelastic period at the design PGA can

be estimated using the much simpler eigenvalue analysis. Using a member stiffness corresponding



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to 50% of the gross section stiffness, the periods obtained from eigenvalue analysis were in good

agreem

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