112

Deep Foundations

113

6.0 Construction and selection of deep foundations

Types of Foundation

The foundations can be divided into two main types, namely: Shallow foundations; and Deep

foundations. Terzaghi categorized foundations into the above categories based on the depth of the

foundation below the existing ground surface and classified the deep foundations as the foundations,

whose depth is more than the width of the foundation. However, classification of the foundations based

on the Terzaghi’s concept serves very little purpose in design and construction of the foundations. For

example, an individual footing foundation, having a depth of embedment more than the width, is

designed and constructed in the same way as a footing, whose depth of embedment is less than the

width. Therefore, for engineering purposes foundations should be classified so that there is a clear

difference between the design and construction of the two types. For this purpose, classification of the

foundations based on the load transfer mechanism to the soil or rock is more appropriate. According to

this classification, as shown in Figure 1.1, foundations with horizontal spreading of the superstructure

load are considered as shallow foundations whereas foundations with vertical load distribution are

classified as deep foundations. Therefore, spread footings, combined footings and raft foundations,

where concentrated forces are distributed laterally, are considered as shallow foundations. Similarly,

piles are the most commonly used type of foundations where vertical distribution of the load takes

place.

Figure 1.1 - Classification of the foundation based on the load transfer mechanism.

Classification of piles

The British Standard Code of Practice for Foundations (BS 8004) places in three categories. These are

as follows.

Large displacement piles - comprise of solid-section piles or hollow-section piles with a closed end,

which are driven or jacked into the ground and thus displace the soil. All types of driven, and driven

and cast-in-place piles come into this category.

Small-displacement piles are also driven or jacked into the ground but have a relatively small cross-

sectional area. They include rolled steel H- or I-sections, and pipe or box sections driven with an open

end such that the soil enters the hollow section. Where these pile types plug with soil during driving

they become large displacement types.

F

F

(b)

Horizontal distribution

of the force

Vertical

distribution

of the force

(a)

114

Replacement piles are formed by first removing the soil by boring using a wide range of drilling

techniques. Concrete may be placed into an unlined or lined hole, or the lining may be withdrawn as

the concrete is placed. Performed elements of timber, concrete, or steel may be placed in drilled holes.

Types of piles in each of these categories can listed as follows.

Large displacement piles (driven types)

1. Timber (round or square section, jointed or continuous).

2. Precast concrete (solid or tubular section in continuous or jointed units).

3. Prestressed concrete (solid or tubular section).

4. Steel tube (driven with closed end).

5. Steel box (driven with closed end).

6. Fluted and tapered steel tube.

7. Jacked-down steel tube with closed end.

8. Jacked-down solid concrete cylinder.

Large displacement piles (driven and cast-in-place types)

1. Steel tube driven and withdrawn after placing concrete.

2. Precast concrete shell filled with concrete.

3. Thin-walled steel shell driven by withdrawable mandrel and then filled with concrete.

Small-displacement piles

1. Precast concrete (tubular section driven with open end).

2. Prestressed concrete (tubular section driven with open end).

3. Steel H-section.

4. Steel tube section (driven with open end and soil removed as required).

5. Steel box section (driven with open end and soil removed as required).

Replacement piles

1. Concrete placed in hole drilled by rotary auger, baling, grabbing, airlift of reverse-circulation

methods (bored and cast-in-place).

2. Tubes placed in hole drilled as above and filled with concrete ass necessary,,

3. Precast concrete units placed in drilled hole.

4. Cement mortar or concrete injected into drilled~hole.:

5. Steel sections placed in drilled hole.

6. Steel tube drilled down.

Composite piles

Numerous types of piles of composite construction may be formed by combining units in each of

the above categories, or by adopting combinations of piles in more than one category. Thus

composite piles of a displacement type can be formed by jointing a timber section to a precast

concrete section, or a precast concrete pile can have an H-section jointed to its lower extremity.

Composite piles consisting of more than one type can be formed by driving a steel or precast

concrete unit at the base of a drilled hole, or by driving a tube and then drilling out the soil and

extending the drill hole to form a bored and cast-in-place pile.

115

In Sri Lanka mostly cast in-situ concrete piles or driven piles are used for pile foundations.

Selection of pile type

Bored piles are constructed by drilling a hole into the ground and filling the hole with concrete

after inserting a reinforcement cage. In comparison, driven piles are constructed by driving a

preformed pile into the ground through application of hammer blows or vibration to the top of

the pile. General factors that should be considered in selecting the type of the pile between

driven piles and bored & cast in-situ piles are discussed in this section while detailed

construction procedures of the two piling methods are discussed in subsequent sections of this

chapter.

Bored piles can be very effectively used when a hard layer is present at shallow depths. The

structural loads can be easily transferred to the hard layer and generally the cast in-situ bored

piles are designed as end bearing piles. Other main advantage of bored piles is its ability to

penetrate minor obstructions, such as boulders, which cannot be penetrated using driven piles.

Such obstructions are commonly found in residual formations in the form of unewathered core

stones commonly referred to as boulders. A driven pile, terminated on a boulder, has a lower

carrying capacity due to the possibility of pile slipping along the face of the boulder and hence,

undergoing large displacement. Therefore, in such situations, use of driven piles is not advisable.

It is very common to find thick weathered or fractured rock above the solid bedrock. There are

certain places, where weathering profile is variable and the rock head is steeply sloping. Under

such situations, it is very important to socket the pile into the bedrock so that the full end bearing

could be mobilized without slipping of the pile toe. Thick weathered zones can be penetrated

using cast in-situ boring techniques and the structural loads can be transferred to the underlying

strong solid rock formations. Compared to driven piles, the ability of the bored piles to penetrate

fractured rock is a tremendous advantage in such situations.

The length of a bored pile can be adjusted easily and in a variable soil or bedrock profile, it is a

definite advantage. Diameter can also be varied and if large diameter piles are used, additional

cost associated with the construction of the pile cap, connecting a group of driven piles, can be

eliminated. The noise and ground vibrations associated with the other driven piling

methodologies are greatly reduced and installation process can be carried out even in a highly

built up area without environmental concerns. Moreover ground heaving associated with

installation of large volume displacement piles is not present with replacement type piles such as

bored and cast in-situ piles. The structural design of the pile should be carried out only

considering the working stresses and reinforcement required for driving and handling stresses

are not needed as in driven piles.

Drilling and concreting is carried out at a certain depth below the ground surface, and in most

cases under a drilling mud. The contamination of the concrete with the drilling slurry, formation

of voids within the pile, necking due to flowing of the sides into the unlined bore, collapsing of

the sides are some of the difficulties associated with bored and cast piles. Moreover, improper

cleaning of the pile bottom can cause considerable reduction in the end bearing capacity of bored

and cast in-situ piles. Formation of a thick hardened layer of bentonite along the sides of the

drilled hole is possible if the bentonite slurry is kept in the borehole for a long period of time.

During concreting the inability of the rising concrete surface to remove the hardened bentonite

may reduce the skin friction along the shaft of bored piles. Therefore, a proper quality

controlling program during the installation process is essential to minimize the defects in the

bored and cast in-situ piles.

As a preformed pile is used, certain concerns associated with bored piles such as necking,

improper cleaning of the borehole, defects due to mixing of concrete and drilling mud etc. are

116

not present with driven piles. When the amount of soil replaced during the installation of the pile

is considered, driven piles fall into small or large volume displacement type. If a pile with a

large volume such as a precast concrete pile is used, the amount of soil displaced during

installation of the pile is large and it is considered as a large volume displacement pile. Whereas

if a pile with a relatively small volume such as a steel pile having a H or I section, is used as a

driven pile the amount of soil displaced during driving is small and such piles are classified as

small volume displacement piles. As a certain amount of soil is displaced during the installation

of a driven pile, the surrounding soil is compacted and thus the soil surrounding the pile is

improved giving rise to increase in the carrying capacity. In comparison, bored piles are

classified as replacement piles and the stress surrounding a borehole for a bored pile is relaxed

resulting reduction of the carrying capacity. Furthermore, the reduction of the skin frictional

capacity due to usage of drilling agents such as bentonite slurry is not present with driven piles.

The structural integrity and the capacity of driven piles are enhanced as the pile is formed under

controlled conditions above the ground surface. But the cast in-situ piles are formed below the

ground surface and in most situations under water making such high level of quality controlling

impossible. Another advantage of the driven piles is the ability to use the piling material

depending on the availability. As an example, when timber is available in abundant, timber piles

can be used for the foundation. Eventhough, noise and vibration generated during driving is very

critical in built up areas, such concerns are not significant in remote areas. Therefore, driven

piles can be used in such situations. If the noise is a concern special techniques such as silent

pile drivers can be used to reduce such environmentally unfriendly effects.

117

Driven displacement piles

Timber piles

In many ways, timber is an ideal material for piling It has a high strength to weight ratio,

it is easy to handle., it is readily cut to length and trimmed after driving, and in favourable

conditions of exposure durable species have an almost indefinite, life. Timber piles used

in their most economical form consist of round untrimmed logs which are driven butt

uppermost. The practice of squaring the timber can be detrimental to its durability since it

removes the outer sapwood which is absorptive to creosote or some other liquid

preservative. The less absorptive heart-wood is thus exposed and instead of a pile being

encased by a thick layer of well-impregnated sapwood, there is only thin layer of treated

timber which can be penetrated by the hooks or slings used in handling the piles, or

stripped off by obstructions in the ground

Timber piles, when situated wholly below ground-water level, are resistant to fungal

decay and have an almost indefinite life. However, the portion above ground-water level

in a structure on land is liable to decay. Although creosote or other preservatives extend

the life of timber in damp or dry conditions they will not prolong its useful life

indefinitely. Therefore it is the usual practice to cut off timber piles just below the lowest

predicted ground-water level and to extend them above this level in concrete. If the

ground-water level is shallow the pile cap can be taken down below the water level.

Bark should be removed from round timbers where these arc to be treated with

preservative. If this is not done the bark reduces the depth of impregnation. Also the bark

should he removed from piles carrying uplift loads by skin friction in case it should

become detached from the trunk, thus causing the latter to slip. Bark need not be removed

from piles carrying compression loads or from fender piles of uncreosoted timber

(hardwoods are not treated because they will not absorb creosote or other liquid

preservatives).

The timber should be straight-grained and free from defects which could impair its

strength and durability. BS 8004 states that a deviation in straightness from the centre-

line of up to 25 mm on a 6 m chord is permitted for round logs but the centre-line of a

sawn timber pile must not deviate by more than 25 mm from a straight line throughout its

length. The Swedish Code SBS-S 23:6 (1968) permits a maximum deviation of 1% of

length between two arbitrarily selected measuring points which must be at least 3 m

apart.

The requirements of BS 8004 of the working stresses in timber piles merely state that

these should not exceed the green permissible stresses given in CP 112 for compression

parallel to the grain for the species and grade of timber being used. The Code suggests

that suitable material will be obtained from stress grades ss and better. Timber piles are

usually in a wet environment when the multiplying factors should be used to convert the

dry properties to the wet conditions. When circulating the working stress on a pile

allowance must he made for bending stresses due to eccentric and lateral loading and to

eccentricity caused by deviations in the straightness and inclination of a pile Allowance

must also be made for reduction in the cross-sectional area due to drilling or notching and

to the taper on a round log.

118

This limitation is applied in order to avoid the risk of damage to a pile by driving it to

some arbitrary ‘set’ as required by a dynamic pile-driving formula and also to avoid a

high concentration of stress at the toe of a pile end bearing on a hard stratum. Damage to

pile during driving is most likely to occur at its head and toe.

The problems of splitting of the heads and unseen ‘brooming’ and splitting of the toes of

timber piles occur when it is necessary to penetrate layers of compact or cemented soils

to reach the desired founding level. This damage can also occur when attempts are made

to drive deeply into dense sands and gravels or into soils containing boulders, in order to

mobilize the required skin-frictional resistance for a given uplift or compressive load.

Judgment is required to assess the soil conditions at a site so as to decide whether or not it

is feasible to drive a timber pile to the depth required for a given load without damage, or

whether it is preferable to reduce the working load to a value which permits a shorter pile

to be used. As an alternative, jetting or pre-boring may be adopted to reduce the amount

of driving required. The temptation to continue hard driving in an attempts to achieve an

arbitrary set for compliance with some dynamic formula must be resisted. Cases have

occurred where the measured set achieved per blow has been due to the crushing and

brooming of the pile toe and not to the deeper penetration required to reach the bearing

stratum.

Damage to a pile can be minimized by reducing: as far as possible the number of hammer

blows necessary to achieve the desired penetration, and also by limiting the height of

drop of the hammer. This necessitates the use of a heavy hammer which should at least

be equal in weight to the weight of the pile for hard driving conditions, and to one-half of

the pile weight for easy driving. The German Code (DIN 183.04) limits the hammer drop

to 2.0 m normally and to 2.5 m exceptionally.

The lightness of timber pile can be an embarrassment when driving groups of piles

through soft clays or silts to a point bearing on rock. Frictional, resistance in the soft

materials can be very low for a few days after driving, and’ the effect of pore pressures

caused by driving adjacent piles in the group may cause the pile already driven to rise out

of the ground due to their own buoyancy relative to that of the soil. The only remedy is to

apply loads to the pile heads until all the piles in the area have been driven.

Heads of timber piles should be protected against splitting during driving by means of a

mild steel hoop slipped over the pile head or screwed to it. A squared pile toe can he

provided where piles are terminated in soft to moderately stiff clays. Where it is

necessary to drive them into dense or hard materials a cast steel point should be provided

(Figure 1.2). As an alternative to a hoop, a cast steel helmet can be fitted to the pile head

during driving. The helmet must be deeply recessed and tapered to permit it to fit well

down over the pile head, allowing space for the insertion of hardwood packing.

119

Figure 1.2 - Typical shoes to be used with timber piles. (Technical specification EI

02G001, US Army Corp of Engineers)

Precast concrete piles

Precast concrete piles have their principal use in marine and river structures, i.e. in

situations where the use of driven-and-cast-in-situ piles is impracticable or uneconomical.

For land structures unjointed precast concrete piles are frequently more costly than

driven-and-cast-in-situ types for two main reasons.

1. Reinforcement must be provided in the precast concrete pile to withstand the

bending and tensile stresses which occur during handling and driving. Once

the pile is in the ground, and if mainly compressive loads are carried, the

majority of this steel is redundant.

2. The precast concrete pile is not readily cut down or extended to suit variations

in the level of the bearing stratum to which the piles are driven.

However, there are many situations for land structures where the precast concrete pile can

be more economical. Where large numbers of piles are to be installed on easy driving

conditions the savings in cost due to the rapidity of driving achieved may outweigh the

cost of the heavier reinforcing steel necessary. Reinforcement may be need in any case to

resist bending stresses due to lateral loads or tensile stresses from uplift loads. Where

high-capacity piles are to be driven to a hard stratum savings in the overall quantity of

concrete compared with cast-in-situ piles can be achieved since higher working stresses

can be used. Where piles are to be driven in sulphate-hearing ground or into aggressive

industrial waste materials, the provisions of sound high-qua1ity dense concrete is

ensured. The problem of varying the length of the pile can be overcome by adopting a

jointed type.

From the above remarks it can be seen that there is still quite a wide range of

employment for the precast concrete pile, particularly for projects where the costs of

establishing a precasting yard can be spread over a large number of piles. The piles can

be designed and manufactured in ordinary reinforced concrete, or in the form of pre-

tensioned or post-tensioned prestressed concrete members. The ordinary reinforced

concrete pile is likely to be preferred for a project requiring a fairly small number of

120

piles, where the cost of establishing a production line for prestressing work on site is not

justifiable and where the site is too far from an established factory to allow the

economical transportation of prestressed units form the factory to the site. Precast

concrete piles in ordinary reinforced concrete are usually square or hexagonal and of

solid cross-section for units of short or moderate length, but for saving weight long piles

are usually manufactured with a hollow interior hexagonal, octagonal or circular sections.

The interiors of the piles can be filled with concrete after driving. This is necessary to

avoid bursting where piles are exposed to severe frost action. Alternatively drainage

holes can be provided to prevent water accumulating in the hollow interior. To avoid

excessive flexibility while handling and driving the usual maximum lengths of square

section piles and the range of working loads applicable to each size are shown in Table

1.1.

Where piles are designed to carry the applied loads mainly in end bearing, e.g., piles

driven through soft clays into medium-dense or dense sands, economies in concrete and

reductions in weight for handling can be achieved by providing the piles with an enlarge

toe.

Table 1.1 - Working loads and maximum lengths for ordinary precast concrete piles of

square section.

Pile size

(mm square)

Range of working Loads

(kN)

Maximum length

(m)

250 200 – 300 12

300 300 – 450 15

350 350 – 600 18

400 450 – 750 21

450 500 – 900 25

BS 8004 requires that piles should be designed to withstand the loads or stresses and to

meet other serviceability requirements during handling, pitching, driving and in service in

accordance with the current standard Code of Practice for the structural use of concrete.

If normal mixes are adopted a 40-grade concrete with a minimum 28-day cube strength of

40 N/mm2 is suitable for hard to very hard driving and for all marine construction. For

normal or easy driving, a 25-grade concrete is suitable. This concrete has a minimum 28-

day cube strength of 25 N/mm2.

To comply with the requirements of BS 8110 precast piles of either ordinary or

prestressed concrete should have nominal cover to the reinforcement as follows.

Exposure conditions

Normal cover for concrete grade of

25 30 40 50 and

over

Buried concrete and concrete continuously

under water

40 mm 30 mm 25 mm 20 mm

Alternative wetting and drying and freezing 50 mm 40 mm 30 mm 25 mm

Exposed to sea water and moorland water

with abrasion

__ __ 60 mm 50 mm

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Concrete cast in shell piles are constructed by driving a steel shell to a required depth by

using a mandrel and filling the shell with concrete after withdrawing the mandrel. Other

three types of concrete piles should be designed considering:

i. Bending stresses developed during handling;

ii. Dynamic stresses developed during driving; and

iii. Stresses due to working loads.

Longitudinal reinforcements are used to carry bending stresses developed during

handling of the precast concrete piles and lateral loads acting on the pile under working

condition. The bending moment diagram of single point handling and the corresponding

handling arrangement for minimum bending moment are given in Figure 1.3(b).

Similarly, the bending moment diagram and the arrangement for minimum bending

moment for double handling point mechanism is given in Figure 1.3(a). Steel stirrups are

used to carry driving stress acting on the pile. However, if the pile is subjected to static

vertical working loads, the reinforcement provided for handling and driving is mostly

redundant under working loads.

Figure 1.3 - Double and single lifting of precast piles: (a) Double lifting, bending moment

diagram and minimum bending moment; and (b) Single lifting, bending moment diagram

and the minimum bending moment.

Prestressed concrete piles have certain advantages over those of ordinary reinforced

concrete. Their principal advantage is in their higher strength to weight ratio, enabling

long slender units to be lifted and driven. However, slenderness is not always

advantageous since a large cross-sectional area may be needed to mobilize sufficient

122

resistance in skin friction and end bearing. The second main advantage is the effect of the

prestressing in closing up cracks caused during handling and driving. This effect,

prestressed pile increased durability which is advantageous in marine structures and

corrosive soils.

The nominal mixes for precast reinforced concrete piles are related to the severity of

driving, and the working stresses appropriate to these mixes are shown in Table 2.6.

For economy in materials, prestressed concrete piles should be made with designed

concrete mixes with a minimum 28-day works cube strength of 40 N/mm2. Metal shoes

are not required at the toes of precast concrete piles where they are driven though soft of

lose soils into dense sands and gravels or firm to stiff clays. A blunt pointed end appears

to be just as effective in achieving the desired penetration in these soils as a more sharply

pointed end and the blunt points is better for maintaining alignment during driving. A

cast-iron or cast-steel shoe fitted to a pointed toe may be used for penetrating rocks or for

splitting cemented soil layers.

During driving of the piles using an impact hammer, a compression stress wave travels

through the pile in the downward directions and reflected at the pile toe to travel upward

direction towards the pile top. If the end resistance at the pile toe is high (fixed end

condition) the reflected wave is compression and on the other hand, low resistance near

the pile toe results in tensile reflection at the pile toe. As a result, driving stresses are

maximum near the pile top and pile toe with reduced driving stresses in the middle

portion of the pile shaft. Therefore, more steel stirrups are provided near the pile top and

pile toe to take up the high driving stresses generated during driving. The requirements of

steel stirrups as specified in BS 8004 are given in Table 1.2 below.

Table 1.2 - The requirements of steel stirrups as specified in BS 8004 for driven precast

piles.

Volume of steel at head

and toe of pile

Volume of steel in

body of pile

Other requirements

0.6% gross volume over

distance of 3 pile width

from each end

0.2% of gross

volume spaced at

not more than ½

pile width

Lapping of shot bars with

main reinforcement to be

arranged to avoid sudden

discontinuity

Steel piles

Steel piles have the advantages of being robust, light to handle, capable of carrying high

compressive loads when driven on to a hard stratum, and capable of being driven hard to

a deep penetration to reach a bearing stratum or to develop a high skin-frictional

resistance, although their cost per metre run is high compared with precast concrete piles.

They can be designed as small displacement piles, which is advantageous in situations

where ground heave and lateral displacement must be avoided. They can be readily cut

123

down and extended where the level of the bearing stratum varies; also the head of a pile

which buck1es during driving can be cut down an re-trimmed for further driving. They

have a good resilience and high resistance to buckling and bending forces. Types of steel

piles include plain tubes, box-sections, H-sections, and tapered and fluted tubes (Mono-

tubes). Hollow section-piles can be driven with open ends. If the base resistance must be

eliminated when driving hollow-section piles to a deep penetration, the soil within the

pile can be cleaned out by grabbing, by augers, by reverse water circulation drilling, or by

airlift. It is not always necessary to fill hollow-section piles with concrete. In normal

undisturbed soil conditions they should have an adequate resistance to corrosion during

the working life of a structure.

Where hollow-section piles are required to carry high compressive loads they may be

driven with a closed end to develop the necessary end-bearing resistance over the pile

base area. Where deep penetrations are required they may be driven with open ends and

with the interior of the pile closed by a stiffened steel plate bulkhead located at a

predetermined height above the toe. An aperture should be provided in the bulkhead for

the release of water, silt or soft clay trapped in the interior during driving. In some

circumstances the soil plug within the pile may itself develop the required base

resistance.

Concrete filling of light-gauge steel tubes is required after driving is completed because

the steel may be torn buckled or may suffer corrosion losses. Piles formed from thin steel

shells driven by means of an internal mandrel, which is withdrawn before filling the

shells with concrete.

The facility of extending steel piles for driving to depths greater than predicated from soil

investigation data has already been mentioned. The practice of welding-on additional

length of pile in the leaders of the piling frame is satisfactory for land structures where

the quality of welding may not be critical. A steel pile supported by the soil can continue

to carry high compressive loads even though the weld is partly fractured by driving

stresses. However, this practice is not desirable for marine structures where the weld

joining the extended pile may be above sea-bed level in a zone subjected to high lateral

forces and corrosive influences.

Bored and Cast In-situ concrete piles

Due to the presence of hard rock layers at relatively shallow depths, bored and cast in-situ

piles are very often used in Sri Lanka. Therefore, the construction procedure of bored and

cast in-situ piles are discussed here.

Replacement piles are installed by first removing the soil by a drilling process and then

construction the pile by placing concrete or some other structural element in the drilled

hole. As mentioned previously, bored piles are constructed by drilling a hole in the

ground and filling it with concrete with or without inserting a reinforcement cage. Since

the borehole in most cases is unlined, there is a possibility of flowing soft soils into the

borehole and forming a ‘necking’ in the pile shaft. Moreover, there could be collapsing of

124

the sides in cohesionless soils and fallen out debris may mix with fresh concrete resulting

in weak spots in the pile shaft. Furthermore, concreting is mostly done underwater

making it impossible to compact fresh concrete. Therefore, special construction

methodologies and precautions had to be followed to ensure a defect free sound bored

and cast in-situ pile. Compared with the construction of shallow foundations,

construction of deep foundation is a challenging task as the construction is carried out at

deeper levels without directly observing it. As a result, indirect quality control measures

should be adopted during the construction process.

In most sites, the ground water table is located at shallow depths and the top soil layers

contain cohesionless soils. Due to the loose soil conditions at the top levels of the ground,

the probability of collapsing of the ground is more. Therefore, it is very common to

install a casing of about 5 to 6m length at the top level of the borehole. If the ground

condition at the lower levels of the subsurface doesn’t contain very loose sandy material,

very often casing of the entire hole is not done and drilling is continued with filling the

hole with bentonite slurry. The cutting through the overburden is usually done by

auguring or chiseling and the cutting debris are removed from the hole using a bucket or

wash boring techniques. Drilling above the water table is usually done using an auger as

shown in Figure 1.4(a) and Figure 1.4(b) shows some auguring tools used for drilling.

(a)

125

(b)

Figure 1.4 – Auguring above the water table: (a) Auguring during drilling; and (b) some

auguring tools used.

Figure 1.5 shows the installation of temporary casing during the drilling process for bored

and cast in-situ piles. After installation of the casing, the center of the casing is checked

as shown in Figure 1.6. During the drilling process, it is very common to use a drilling

fluid, such as bentonite slurry, to keep the sides of the borehole stable. The borehole is

filled with drilling fluid, when the borehole reaches the ground water table. For this

purpose, a bentonite reservoir is formed either surrounding the pile bore or away from the

pile bore location. Figure 1.7 shows a bentonite reservoir surrounding the pilebore and

Figure 1.8 shows a bentonite reservoir away from the pilebore location.

Figure 1.5 – Installation of a temporary casing.

126

Figure 1.6 – Checking the location of the center of the pile.

Figure 1.7 – Bentonite reservoir surrounding the pile bore.

127

Figure 1.8 – Bentonite reservoir away from the location of the pilebore.

Use of Bentonite as a drilling mud

Two types of natural bentonite exist: swelling bentonite which is also called sodium

bentonite and non-swelling bentonite or calcium bentonite. Sodium bentonite expands as

it can absorb several times its dry weight of water. It is mostly used in drilling mud in the

oil and gas well drilling industries as it exhibits low filter loss. However, non-swelling (or

low-swelling) bentonite has much higher filter or fluid loss than swelling sodium

bentonite and hence, it is not effective as a drilling fluid. As it is commonly accepted, the

drilling mud should perform or facilitate following tasks:

i. Remove cuttings produced by the bit at the bottom of the hole and carry them

to the surface;

ii. Lubricate and cool the drill bit during operation, as friction causes high

temperatures down-hole that can limit tool life and performance;

iii. Maintain hydrostatic equilibrium so that water from the surrounding soil do

not enter the borehole causing the wall to flow, kink and blow out. This is

achieved by adjusting the mud weight (density);

iv. Build a filter cake (or skin) on the wall of the drilled hole, preventing fluid

loss by mud invasion of penetrated formations; and

Pump used to circulate

bentonite

128

v. Support and prevent caving of the wall of the hole.

Typically if 3% or more of bentonite powder is dispersed in water, a viscous slurry is

formed which is thick when allowed to stand but becomes thin when agitated. This

phenomenon is referred to as thixotropy.

Bentonite slurry provides the stability to the borehole walls by two main actions: (i)

Formation of a filter skin termed “cake” at the interface of the slurry and the walls of the

excavated hole; and (ii) higher lateral pressure of the dense slurry pushing against the

filter skin and the walls of the excavated hole. The concreting of the hole should be done

in such a way to displace the slurry in the hole with the fresh concrete. However, if the

slurry full of hole is kept for a long period of time, a thicker and harder “cake” will be

formed on the internal walls of the borehole. If the soil surrounding the pile shaft is

permeable, the water in the bentonite slurry may seep into the surrounding area forming a

thicker filter cake. Some researchers have shown that it is possible to form a thin cake of

few millimeters even in clayey soils, which is quite impermeable. The formation of filter

cake in clayey soils to electrical forces or chemical reaction of bentonite suspension on

the wall of the borehole. It is argued that if the shear strength of the filter cake formed is

more than that of the fluid concrete, it cannot be scoured by the rising concrete surface

during concreting and may be left in place resulting in degradation of the development of

skin friction.

It is believed that the formation of the major portion of the filter cake, and hence the

reduction of the major portion of the skin friction capacity, takes place within first few

hours of the construction time and further increase in construction time have minor effect

on the reduction of the skin friction capacity. The formation of the filter cake, which

reduces the development of skin friction, takes place at a higher rate within first few

hours between the end of drilling and concreting. The delay time between the end of

drilling and concreting should be minimized to reduce the effects of the filter cake on the

development of skin friction in bored and cast in-situ piles. .

Formation of the filter cake takes place as the water in the bentonite slurry seeps to the

surrounding area in sandy soils or chemical action between bentonite slurry and the

surrounding clayey soils. Within the rock socketed length of the pile, seeping of the water

takes place through the cracks in the rock mass. If the cracks are open and filled with

high permeable debris, large quantity of water may seep into the surrounding area and

formation of the filter cake may be enhanced. However, if the rock mass surrounding the

socketed length is impervious, the filter cake formed may be limited to a very thin layer.

If the bottom of the borehole is cleaned immediately before concreting, there is a high

probability that the thin layer of the filter cake formed within the rock socket may be

scraped off.

Drilling below the water table

The drilling below the water table can be carried out using rotary drilling or percussion

drilling. If a rotary drilling method is used, a drilling bucket, as shown in Figure 1.9, is

used to remove material from the pile bore and it is emptied, as shown in Figure 1.10. On

the other hand, a chisel is lifted and dropped to loosen the material in the percussion

drilling and then wash boring is used to remove the debris from the pile bore. Figure 1.11

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shows rigs used in the percussion drilling.

Figure 1.9 – Removing debris from the pile bore.

Figure 1.10 – emptying the drilling bucket.

130

Figure 1.11 – Percussion drilling rigs.

Figure 1.12 – Rotary drilling tool used to drill through rock.

131

Termination of pile bore

During the drilling process for the piles, rotary drilling or percussion drilling techniques

are used to drill through the rock. Since the coring through the rock is very rarely done,

the residue coming out consists of very small rock particles, which hardly gives any

indication of the quality of the bedrock. Therefore, the quality of the bedrock should be

established by some other means.

It is not uncommon in Sri Lanka to find sites with fairly thick weathered rock layers

overlying the sound bedrock. In this type of sites, very often large variation in pile

lengths are reported within very short distances. Due to the steep dip angle of the bedrock

and the highly fractured nature of the bedrock, a pile termination criterion plays a special

significance in this type of sites. A detail site investigation including the investigation of

the bedrock is a must for these sites to design a suitable pile foundation and to plan the

construction phase of the foundation. Another weakness in the site investigation

procedure adopted in Sri Lanka is the lack of coordination between the site investigation

firm and the client and/or the consultant. If the site investigation firm informs the site

conditions, for example the variation of the bedrock profile and the quality of the bedrock

at the site, to the client and/or the consultant during the field investigation phase then, the

site investigation program can be modified to suit the site conditions.

If the establishment of bedrock was not properly done during the site investigation

process, identification of the bedrock in a variable bedrock profile is highly questionable.

In a site, where the bedrock elevation highly varies across the site, during the site

investigation stage rock drilling should be carried out at reasonable number of points

across the site to establish the bedrock level with a relatively high RQD. Such

investigation will not only give more data needed for the design of the pile foundation but

also will provide very important information needed for planning the construction

process as well.

In a typical site with varying bedrock profile, it is very difficult to identify the bedrock

and estimate the pile socketing length during drilling for the piles. Some piling

contractors use highly subjective criteria such as penetration rate of the drilling tool as a

guide to establish the bedrock level. However, socketing length and the termination

criterion of the piles based on the rate of penetration of drilling tools could be highly

misleading as the drilling through the bedrock could give high and low penetration rates

depending on the weathered nature of the fractures in the upper part of the bedrock and

the quality of the cutting tool.

The pile termination criterion, for a site with varying bedrock profile preferably should be

done after installation of a test pile near a location of a borehole used for field

investigation. The information obtained during drilling for the test pile and the load test

result obtained from the test pile should be used for determination of the carrying

capacity of piles in the site and in deciding the termination criterion to be used for

installation of the production piles. It should noted here that a test pile should be loaded

upto twice the working load as specified in section 6.2 of ICTAD/DEV/15, not upto 1.5

times the working load as testing of a working pile.

Once the termination criteria for the site are established, the drilling process should start

from one side of the site and proceed forward. Level of the bedrock should be marked on

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a site map and preferably contours of the pile tip elevation should be plotted. The contour

map should be updated as the drilling progress and the pile tip elevation contours should

always be compared with the elevation of the bedrock established during the initial site

investigation process. Large variation of the pile tip elevation of a new pile from the

elevation shown by the contours should be carefully studied. For example, if the bedrock

is encountered at a higher elevation than the expected elevation of the bedrock from the

already established bedrock contours, drilling should be done to make sure that the pile

has not hit a core stone in the weathered rock layer.

Cleaning of the Borehole before Concreting

This is another very important aspect, which is not given due attention, during

construction of the piles. If the pile bottom is not properly cleaned before concreting is

done, there could be a layer of waste material present between the bottom of the pile and

the bedrock. As this material consists of unconsolidated loose debris, when the pile

bottom is loaded, it will undergo large settlement. The debris that is present may consist

of:

i. Granular material from the drilling operation through rock and soil, which is

in suspension with the drilling mud, may settle to the bottom of the borehole;

ii. Small block-like portions of soil and rock from the unlined wall of the

borehole may dislodge and fall down to the bottom of the borehole; and

iii. Ground water percolated through the pervious silty and sandy layers may

transport and deposit significant amount of sandy and silty material at the

bottom of the unlined borehole.

Even if the concreting procedure is methodical to give a defect free pile shaft, the

presence of the loose material below the pile bottom severely hamper the load carrying

capacity of the pile due to large settlement it undergoes.

Through the surveys carried out in Sri Lanka it is found that about 5% of the piles are not

according to the specifications and are categorized as ‘defective’. The analysis of the

load-settlement curves and the site conditions of the ‘defective’ piles indicated that the

piles have undergone large plunging type settlement under very small end bearing

resistance due to the presence of a relatively soft layer below the bottom of the pile.

Therefore, the most probable reason for presence of a soft layer beneath the pile toe is the

improper cleaning of the bottom of the borehole before concreting.

Concreting of Cast in-situ Bored Piles

The borehole may be dry, partially or completely filled with fluid before concreting.

Concreting under dry conditions should be done from dropping concrete from the ground

surface so that concrete ‘free falls’ onto the base of the borehole. A hoper or a guide

trunk should be used at the ground surface level to avoid contamination of the concrete

with soil near the ground surface level. The mix design of the concrete should be done to

produce a workable mix, which is self compacting without segregation. However,

segregation of fresh concrete may take place when the falling concrete hits the

reinforcement cage. To reduce the segregation of concrete this way, some contractors

133

sprinkle cement powder on to the reinforcement cage before concreting.

Concreting a borehole partially or completely filled with fluid is a difficult task and

requires careful planning and supervision to construct a defect free pile. Since it is not

possible to observe the actual concreting process taking place down the borehole, some

indirect quality controlling measures should be adopted to ensure defect free pile. The

contractor, in consultation with the consultant to the project, should device a suitable

quality control program prior to the beginning of the piling process. In devising such

quality control program, due consideration should be given to the subsurface conditions

of the site. It is very often observed that the piling contractors don’t change their

construction process to suite the subsurface condition.

The concreting of the pile under water should be carried out using a tremie pipe. The

tremie pipe should be watertight and the interior surface should be free from any

projections for unhindered passage of concrete through it. Typically 125mm to 200 mm

diameter tremie pipes are used to concrete bored piles in Sri Lanka. Usually larger

diameter pipes are used to concrete large diameter piles and/or concrete with large

aggregates. It should be reiterated here that before the commencement of concreting,

drilling mud at the bottom of the borehole should be checked for contamination.

The tremie pipe is assembled inside the borehole, which is full of bentonite slurry. The

funnel (or hoper) is attached at the top of the assembled tremie pipe, which is long

enough to reach the pile bottom as shown in Figure 1.13(a) A plug is placed at the bottom

of the hoper and a small volume of suitable buoyant material is placed between the

bottom of the fresh concrete in the funnel and slurry in the tremie pipe as shown in Figure

1.13(b). The purpose of the buoyant material is to keep the first batch of concrete mixing

with the slurry in the tremie pipe. Otherwise, during the falling of the first batch of

concrete through the tremie tube, washing of concrete and mixing it with bentonite slurry

may take place significantly weakening the concrete. The hopper is filled with concrete

with the removable plug placed at the bottom of the hopper. Then, the plug is jerked out

allowing fresh concrete to shoot down under its own weight to the bottom of the

borehole. Concrete rapidly moving down the tremie pipe may push the drilling mud in

the pipe through the bottom as shown in Figure 1.13(c). Thus the first charge of concrete

is placed and the bottom of the tremie is immersed in fresh concrete to create a sealed

environment inside the tremie from the drilling mud outside. There are two potential

problems associated with initial charging of tremie with concrete:

(i) Segregation of concrete during placement and;

(ii) Entrapment of air inside the tremie pipe.

To avoid these problems, the tremie should be filled slowly after placing the initial

charge. During the time period, from initial charging of the pile to end of concreting, the

bottom of the tremie pipe should be always kept below the top surface of the concrete

inside the borehole. The depth of embedment of the tremie pipe in the borehole should be

about 1.5m to 3.0m and higher depth of embedment should be used for concreting large

diameter piles.

134

Figure 1.13 - Concreting a borehole using tremie pipe: (a) Tremie is assembled in the

borehole; (b) A plug is placed at the bottom of the hopper and filled with concrete; (c)

Plug is removed and concrete moving through the tremie; and (d) Concreting continued

with bottom of tremie pipe immersed in fresh concrete.

The specific gravity of the drilling mud may go up with the degree of contamination of

the drilling mud with silt and other debris and the specific gravity of the drilling mud

should be less than 1.25 before the beginning of the concreting process. If the drilling

mud is contaminated with drilling debris and other substances, additional recycling or

substitution of the suspension is necessary so that the flow of fresh concrete can readily

replace the drilling mud at the bottom of the borehole. Concreting should be done in a

continuous operation without any interruptions. Therefore, the site engineer should make

necessary arrangements for continuous supply of concrete without delay. A contingency

plan should also in place to supply concrete if delay in the expected supply of concrete

happened due to some unforeseen reasons.

It is observed at most of the sites, that the tremie pipe is lifted up and lowered rapidly to

facilitate rapid flow of concrete. Since rapid lifting and lowering of the tremie causes the

mixing of drilling mud and the concrete within a certain zone surrounding the tremie,

such practice should be minimized or such movement should be limited to a small height.

Due to some reason if the tremie bottom is taken out of the fresh concrete, placement of

concrete should be stopped and the following procedure should be adopted in

recommencing the concreting process.

135

The tremie should be gently lowered on to the surface of the

previously placed concrete with very little penetration. The tremie

should be filled with high slump concrete with higher cement content

and a new initial charging of the tremie should be done to displace the

laitance/scum at the top of the old concrete surface with fresh concrete.

The tremie should be pushed further slowly making fresh concrete

sweep away laitance/scum in its way.

However, if there is any delay in recommencing the concreting of the borehole, the above

procedure may not be applied as replacement of laitance/scum of set or partially set

concrete cannot be effectively carried out. In such situations, a new pile fully or partially

replacing the problematic pile should be introduced.

Withdrawing the casing is another important process that has to be done during the

concreting process. The rate of withdrawing the casing is the governing factor. If casing

is withdrawn too fast, the

Minipiles and micropiles

Minipiles are defined in CIRIA report PGI(2.10) as piles having a diameter of less than 300

mm, with working loads in the range of 50 to 500kN. The term “micro-pile is given to

those in the lower range of diameter. They can be installed by a variety of methods.

Some of these are:

i. Driving small-diameter steel tubes followed by injection of grout with or

without withdrawal of the tubes;

ii. Driving thin wall shells in steel or reinforced concrete which are Oiled with

concrete and left in place;

iii. Drilling holes by rotary auger, continuous flight auger, or percussion

equipment followed by placing a reinforcing cage and in-situ concrete in a

manner similar to conventional bore pile construction;

iv. Jacking-down steel tubes, steel box-sections. or precast concrete sections. The

sections may be jointed by sleeving or dowelling.

The principal use of minipiles is for installation in conditions of low headroom such as

underpinning work or for replacement of floors of buildings damaged by subsidence.

Factors governing choice of type of pile

The advantages and disadvantages of the various forms of pile described in 2.2 to 2.5

affect the choice of pile for any particular foundation project and these are summarized as

follows:

Driven displacement piles

Advantages

1. Material forming pile can be inspected for quality and soundness before driving.

2. Not liable to ‘squeezing’ or ‘necking’.

3. Construction operations not affected by ground water.

136

4. Projection above ground level advantageous to marine structures.

5. Can be driven in very long lengths.

6. Can he designed to withstand high bending and tensile stresses.

Disadvantages

1. Unjointed types cannot readily be varied in length to suit varying level of bearing

stratum.

2. May break during driving, necessitating replacement piles.

3. May suffer unseen damage which reduces carrying capacity.

4. Uneconomical if cross-section is governed by stresses due to handling and driving

rather than by compressive, tensile, or bending stresses caused by working conditions.

5. Noise and vibration due to driving may be unacceptable.

6. Displacement of soil during driving may lift adjacent piles or damage adjacent

structures.

7. End enlargements, if provided, destroy or reduce skin friction over shaft length.

Driven-and-cast-in-place displacement piles

Advantages

1. Length can easily be adjusted to suit varying level of beating stratum

2. Driving tube driven with closed end to exclude ground water

3. Enlarged base possible

4. Formation of enlarged base does not destroy or reduce shaft skin friction

5. Material in pile not governed by handling or driving stresses

6. Noise and vibration can be reduced in some types by driving with internal drop-

hammer

Disadvantages

1. Concrete in shaft liable to be defective in soft squeezing soils or in conditions of

artesian water flow where withdrawable-tube types are used.

2. Concrete cannot be inspected after installation.

3. Length of some types limited by capacity of piling rig to pull out driving tube

4. Displacement may damage fresh concrete in adjacent piles or lift these piles or

damage adjacent structures. . 5. Noise and vibration due to driving may be unacceptable

6. Cannot be used in river or marine structures without special adaptation.

7. Cannot be driven with very large diameters.

8. End enlargements arc of limited size in dense or very stiff soils.

9. When light steel sleeves arc used in conjunction with withdrawable driving tube, skin

friction on shaft will be dest toyed or reduced.

137

Bored-and-Cast-in-Place replacement piles

Advantages

1. Length can readily he varied to suit variation in level of bearing stratum.

2. Soil or rock removed during boring can be inspected comparison with site

investigation data

3. In-situ loading tests Can be made in large-diameter pile boreholes, or penetration tests

made in small boreholes.

4. Very large (up to 7.3m diameter) bases can be formed in favourable ground.

5. Drilling tools can break up boulders or other obstructions which cannot be penetrated

by any form of displacement pile.

6. Material forming pile is not governed by handling or driving stresses.

7. Can be installed in very long lengths.

8. Can be installed without appreciable noise or vibration.

9. No ground heave.

10. Can be installed in conditions of low headroom.

Disadvantages

1. Concrete in shaft liable to squeezing or necking in soft soils where conventional types

are used.

2. Special techniques needed for concreting in water-bearing soils.

3. Concrete cannot be inspected after installation.

4. Enlarged bases cannot be formed in cohesionless soils.

5. Cannot be extended above ground level without special adaptation.

6. Low end-bearing resistance in cohesionless soils due to loosening by conventional

drilling operations

7. Drilling a number of piles in group can cause loss of ground and settlement of

adjacent structures.

138

Design of Piles

139

7.0 Design of Piles

Design criteria

Similar to a shallow foundation, failure of a structurally intact pile can be caused due to

two reasons: (i) shear failure of the soil surrounding the pile and; (ii) excessive settlement

of the foundation. Therefore, the task of the foundation designer is to find out an

economical pile to carry the working load with a low probability of shear failure while

keeping the resulting settlement to within allowable limits. In designing a single pile

against shear failure, it is customary to estimate the maximum load that can be applied to

a pile without causing shear failure, generally referred to as the ultimate carrying

capacity.

As in the case of shallow footings, two design approaches, (1) Allowable Stress Design

(ASD) method and (2) Load Resistance Factor Design (LRFD) method are available for

piles. The following sections will mostly elaborate the ASD method. The allowable stress

design (ASD) requires the following conditions:

Allowable loads

alln QFSP / (6.1 )

where

nP = ultimate resistance of pile

allQ = allowable design load

FS = factor of safety

The ultimate working load that can be applied to a given pile depends on the resistance

that the pile can produce in terms of side friction and point bearing (Figure 6.2). As the

pile is loaded at the pile top, the pile tends to move in the downward direction relative to

the surrounding soil. Therefore, the surrounding soil offers resistive force against that

relative movement. Hence the expression for the allowable load Pa on a pile would take

the following form:

FS

PPQ

supu

all

(6.3)

where

Ppu = ultimate point capacity

Psu = ultimate side friction

Determination of the ultimate carrying capacity of piles

There are mainly two different methods available to estimate the ultimate carrying

capacity of piles: i. static methods, and ii. dynamic methods.

Static methods can be further divided into following methods:

a. Using strength parameters of soil and/or rock;

b. Using empirical correlations and in-situ test results; and

c. Using static pile load test results.

140

Dynamic methods can also be sub-divided into following methods:

a. Using pile driving equations;

b. Using the wave equation method; and

c. Dynamic testing of the piles.

Figure 2.1 Load carrying mechanism of piles.

Static methods of estimation of the ultimate carrying capacity of piles

The ultimate carrying capacity of piles is the maximum load that can be applied on the

pile without causing shear failure of the surrounding soil both along the pile shaft and at

the pile bottom. As the skin friction may not be uniform along the pile shaft, the skin

friction is estimated by adding the skin friction along the pile shaft.

siupu PPP ,

However, it is observed that the deformation required to develop the ultimate point

bearing capacity is much higher compared to the deformation required to develop the

ultimate skin frictional capacity. Therefore, some define the ultimate carrying capacity of

the pile as summation of the ultimate skin friction and the developed end bearing capacity

when the skin friction reaches the ultimate value.

usipu PPP ,

The total pullout resistance of the pile may be estimated using the following Equation:

Friction along the pile

shaft (skin friction)

Resistance at the pile point

141

pusiu WPT ,

uP = Ultimate (maximum) pile capacity in compression-usually defined as that load

producing a large penetration rate in a load test

Tu = Ultimate pullout capacity

Pp, u = ultimate pile tip capacity – seldom occurs simultaneously with ultimate skin

resistance capacity usiP , : neglect for floating piles (which depends only on skin

resistance)

pP = tip capacity developed simultaneously with usiP , : neglect for “floating piles”

siP = skin resistance developed simultaneously with ultimate tip resistance Pp, u

: neglect for point bearing piles

usiP , = ultimate skin resistance developing simultaneously with some tip

resistance Pp

Wp = weight of the pile being pulled

= summation process over I soil layers making up the soil profile over length

of pile shaft embedment

The ultimate capacity of a pile can be generally written as:

bbusuu WPPP

suP = ultimate shaft resistance

buP = ultimate base resistance

Wb = weight of the pile

Skin Friction

Development of skin friction in piles

A pile, which is in contact with the soil along its shaft, is loaded as shown in Figure

2.2(a). Due to the higher stiffness of the pile material relative to that of the soil, as the

load on the pile is applied, the pile tends to move in the downward direction relative to

the surrounding soil. This is similar to the situation, where two solid objects in contact

with each other, one of the objects tries to move relative to the other object. Naturally, a

resistive force is developed between the two objects to resist that attempted movement.

If two soil and pile elements in contact with each other are considered, the pile element

tends to move in the downward direction relative to the surrounding soil element as

shown in Figure 2.2(b). An imaginary space is created between the pile and the soil

elements, in reality they are in contact with each other. As the pile element tends to move

142

in the downward direction relative to the surrounding soil element, the soil element also

moves with it and the downward moving soil element applies an upward resistive force

(fs) on the pile element, as shown in Figure 2.2(b). The pile element apply an equal an

opposite downward force on the soil element.

As the downward displacement of the pile element increases, the resistive force

developed between the pile and the soil elements is increased as shown in Figure 2.2(c).

However, that resistive force cannot increase indefinitely. The resistive force developed

reaches a maximum, commonly referred to as the ultimate skin friction (fs,u). The

relationship between the skin frictional force and the downward deflection of the pile

element can be approximated as shown in Figure 2.2(c). The downward displacement

required to mobilize the ultimate skin friction resistance is relatively low and is in the

rage of 5 – 10mm. After the ultimate skin friction is mobilized, the pile and the soil

elements start to slip.

Figure 2.2 – Development of skin friction on pile.

Load Transfer Curves.

The axial force variation in the pile with the depth is referred to as the load transfer curve.

If the pile is not subjected to negative skin friction, the axial force in the pile is maximum

at the pile top and is equal to the applied force. Considering the static equilibrium of the

section of the pile upto a depth z, the following Equation could be written:

fs

fs

Pile element Soil element

Imaginary separation between

pile and soil elements

(a)

(b)

(c) Relative displacement of

the pile element ()

Skin friction developed (fs)

fs,u

o

143

1sazat fPP

From the above Equation it is clear that if the skin friction is acting in the upward

direction, the axial force decreases with the depth.

Considering the equilibrium of the small element of length dz shown in Figure, following

equilibrium equation could be written:

sdzzaaz dfPP )(

Re-arranging the terms in the above Equation,

)( dzzaazs PPdf

The skin friction at a given section is equal to the difference in the axial force at that

section.

Figure 2.2 - Axial force along the pile axis.

fs1

dfs

(a) (b)

z

dz

Pa Pat

Paz

Pa(z+dz)

144

Figure 2.4 load transfer curves obtained by increasing the load acting on the top of a pile

in clayey soil.

Figure 2.5 load transfer curves obtained by increasing the load acting on the top of a pile

in sandy soil.

145

If the load transfer curve is vertical at a given section, it indicates that the skin friction in

that section is zero. Figures 2.4 and 2.5 are load transfer curves obtained by varying the

force acting at the top of piles and clay and sand respectively. Careful observation of

Figure 2.4 clearly indicates that when the force on the pile is increased from 300 kips, the

shape of the curves do not change significantly but the curves are shifted to the right (i.e.

the difference in the axial force between any two sections of the pile shaft does not

change significantly). This is due to reaching of the skin friction to the ultimate value.

Once the skin friction reaches the ultimate value along the entire pile shaft, the additional

load increased at the pile top directly increases the end bearing at the pile bottom. The

other point to note is the mobilization of the end bearing capacity. Initially only a very

small portion of the end bearing capacity is mobilized. But as the skin friction reaches the

ultimate capacity, the end bearing resistance increases significantly, finally reaching a

situation where load increment at the pile top is causing equal increment in the end

bearing resistance. From this it could be concluded that the initial load increments on

piles are taken up by mobilizing skin friction and very minimal end bearing is mobilized.

However, closer to failure load increments are entirely taken by increase in end bearing.

The load transfer curves shown in the Figure 2.5 also confirms the above facts and clearly

indicates that for HP 14 x 89 pile the shape of the load transfer curve do not change

significantly after 100 kips load at the pile top. This indicates that the skin friction has

reached the ultimate value by then. However, the end bearing capacity increases beyond

that. It was earlier mentioned that the skin friction reaches the ultimate capacity at low

deformation levels whereas the end bearing reaches the ultimate capacity at a very high

deformation levels.

Estimation of the ultimate skin frictional capacity of piles

Using soil strength parameters

As explained earlier, the development of the skin friction is due to the movement of the

pile shaft relative to the surrounding soil. At any level of the pile shaft a normal force (fn)

is applied on the pile shaft from the surrounding soil as shown in Figure 2.6.

146

Figure 2.6 – Development of frictional resistance.

Assuming that the soil element in contact with the pile is also solid, theory of friction

between two solid objects can be used to find out the frictional resistance developed.

From fundamental concepts in frictional resistance between two solid objects, the

frictional resistance (ffr) developed can be expressed as:

tannnfr fff

Where,

- Coefficient of friction between the two objects; and

- Friction angle between the two objects.

At a given location of the pile shaft, the normal force fn may be assumed to be a constant.

The friction angle, , is not a constant and it increases with the relative displacement

between the two objects. As the displacement of the pile element, as shown in Figure 2.2,

increases the frictional resistance, ffr, increases upto the ultimate value, fufr. The maximum

friction angle between the soil and the pile element is taken as a, which is generally a

function of the angle of internal friction of the soil. Therefore, the ultimate frictional

resistance (fufr) may be expressed as:

annufr fff tan

Dividing both sides of the above Equation by the surface area of the element, As:

anufr tan

Where

fs

fs

Pile element Soil element

Imaginary separation between

pile and soil elements

(a)

(b)

fn

147

ufr - Ultimate unit frictional resistance

n - Stress normal to the pile – soil interface

In addition to the frictional resistance developed between the pile and the soil, a unit

adhesive resistance, ca, may also be developed if the soil is cohesive. Therefore, the total

ultimate skin frictional resistance, us, may be expressed as:

aufrus c

By substitution,

anaus c tan

It is generally observed that the:

'

vsn K

Where Ks – coefficient of lateral earth pressure

v/ - effective vertical stress at the level considered.

The lateral earth pressure coefficient is a function of the soil type, stress history and the

amount of disturbance caused to the surrounding by the pile installation process.

Thus,

asvaus Kc tan

and

dzCP us

L

us 0

dzKcCP asva

L

us tan0

where

C = pile perimeter

L = length of pile shaft

Estimation of the skin friction in clayey soils

In clayey soils, the undrained condition is critical. Therefore, the ultimate skin frictional

resistance should be estimated using undrained strength parameters.

For clayey soils under undrained condition = 0 and hence, a = 0. Therefore,

aus c

It was found that:

uus c

Where

cu - Undrained cohesion of clay.

Different researches have suggested different relationships for and one of the very

widely used simple relationship is given below:

148

Figure2.7 – Relationship between and undrained strength (Su)

Example

Estimate the ultimate skin frictional resistance of the 400 mm x 400 mm, 9m long precast

concrete pile driven into clyey soil having undrained cohesion of 50 kPa.

Solution

From Figure 2.7, = 0.9 (Using the relationship proposed by Bowles (1996))

509.0 xcuus = 45 kPa

dzCP us

L

us 0

cu = 50 kPa

400 mm x 400 mm driven

pile

9 m

Bowles (1996)

149

Since the us is constant with the depth, the total ultimate skin frictional capacity may be

estimated by multiplying us by the total surface area of the pile shaft.

Therefore,

94.04.0245 xxxPus = 648 kN.

Ultimate skin frictional capacity is 648 kN.

Exercise

Estimate the ultimate skin frictional capacity of the 9m long, 400 mm x 400 mm square

pile driven into a subsurface consisting of two clay layers having undrained cohesion of

25 kPa and 50 kPa as shown below. Estimate the ultimate skin frictional capacity of the

pile.

Estimation of the skin friction in clayey soils

The general skin friction capacity can be expressed by the Equation derived earlier:

asvaus Kc tan

For sandy soils, ca = 0 and hence the above Equation is simplified to:

asvus K tan

It is a very difficult task to accurately estimate the value of the lateral earth pressure

coefficient as it may depend on the type of soil, method of installation etc. As a result,

there are different methodologies proposed by various researches to estimate the

coefficient of lateral earth pressure closer to the pile. In this course, the -method

proposed by Burland (1972) is discussed.

-method

Burland (1972) made the following assumptions:

cu = 25 kPa

400 mm x 400 mm driven

pile

9 m

cu = 50 kPa

4 m

150

1. Due to remolding adjacent to the pile, the effective stress cohesion intercept

reduced to zero;

2. The horizontal stress acting on the pile after dissipation of pore pressure is at least

equal to the horizontal stress prior to the installation of the pile (K0 condition); and

3. The major shear distortion during pile loading is confined to a relatively thin zone

around the pile shaft, and drainage of this thin zone either occurs rapidly during loading

or has already

occurred in the delay between driving and loading.

Considering the above assumptions, the skin friction can be expressed as:

aovus K tan

Where

Ko - Coefficient of lateral earth pressure at rest and for normally consolidated

soils it is equal to (1 - sin).

Substituting Kotan a = , the ultimate skin frictional capacity is given by:

vus

A particular attractive feature of this method is that if we use '

0 and 1 SinK

then the range of is from 0.27 to 0.30 in the pracital (range of 25° to 45°). This method

is more of then used with piles driven in cohesionless soil (when 0 ,0' ).

Variation of the effective overburden pressure (v) closer to the pile

In undisturbed ground, the effective overburden pressure increases with the depth, as the

weight of the soil above a certain level increases with the depth. However, this situation

changes closer to a pile when it is installed in the ground. Closer investigation of the skin

friction Equation given above reveals that at higher depths, if the effective overburden

pressure increases with the depth, the skin friction capacity should be very high. But

researches have found that the skin friction of piles do not increase without bounds with

the depth. The results of skin friction variation in sandy soils measured using model piles

are presented in Figure 2.8. It is clear from the variation of the skin friction with the

depth that the skin friction varies upto a certain depth but beyond that it remains constant.

151

Figure 2.8 – Variation of the skin friction with the depth (Vesic, 1967)

Based on the results of the research it is concluded that the effective vertical overburden

pressure closer to the pile is not similar to the vertical effective overburden pressure

under in-situ undisturbed conditions. The presence of the pile tends to change the stress

conditions closer to the pile. The pile provides some arching action and tends to reduce

the overburden pressure beyond a certain critical depth as shown in Figure 2.9.

Zc

W. T.

L

(a) (b)

wetdw+(Zc-dw)(sat- w)

wetdw+(L-Zc)(sat- w)

152

Figure 2.9 – (a) Vertical effective vertical stress distribution closer to the pile; and (b)

Vertical effective vertical stress distribution away from the pile.

It should be noted here that the skin friction developed on piles in sand varies with the

depth and the total resistance should be estimated considering such variations. The

relationship shown in Figure 2.10 can be used to determine the critical depth of a pile.

Figure 2.10 – Critical depth (zc) / pile diameter vs friction angle of the soil.

The angle of internal friction 1, prior to the installation of the pile, should be modified as

follows before using with Figure 2.10.

For driven piles;

104

31 o

For bored piles;

31 oo

Example I

A driven 400mm square, 9.0m long pile is installed in sandy soil layer having angle of

internal friction = 32o and cohesion c = 0. The water table is present at 1m below the

ground surface and the unit weight of soil above and below the water table are 16 kN/m3

and 17 kN/m3 respectively.

i. Determine the skin friction distribution along the pile; and

ii. Estimate the total ultimate skin frictional resistance on the pile.

153

Solution

i.

Equivalent diameter (d) of the pile can be determined by finding the diameter of the pile

having the same cross sectional area as the square pile.

mmxB

d 45142

Modification of the friction angle;

ooo x 3410324

310

4

31

From Figure 2.10, Zc/d = 6, therefore, the critical depth Zc = 2.70m.

Effective overburden pressure at the water table level = 116x = 16 kPa

Effective overburden pressure at the critical depth level = 0.170.216 wsat

= 28.2 kPa.

The angle adhesion, a = 0.75 = 24o

21.0tansin1 a

Ultimate skin friction:

At the water table level = 1621.0 xv = 3.36 kPa

At the critical depth level = 2.2821.0 xv = 5.92 kPa.

Since the vertical effective overburden pressure closer to the pile remains constant

beyond the critical depth and is also constant, the skin friction is constant below the

critical depth level.

wet= 16 kN/m3

Sat= 17 kN/m3

1 m

W. T

8 m

400 mm x 400 mm driven pile

= 32o

154

The skin friction distribution can be graphically shown as below:

ii.

The total ultimate skin friction can be estimated by determining the area of the skin friction variation with the depth.

Total skin friction upto the WT = 0.136.32

1x = 1.68 kN

Total SF from 1.0 to 2.70m = 0.170.292.536.32

1 = 7.79 kN

Total SF from 2.70 to 9.0m = 7.20.992.5 x = 37.3 kN

Total SF = 46.8 kN Example II

A 800 mm diameter, 22 m long bored pile is installed through the subsurface shown in

the following Figure. The water table is present at 2 m below the ground surface and the

unit weight and the strength properties of the different layers are also given in the same

Figure.

i. Determine the skin friction distribution along the pile; and

ii. Estimate the total ultimate skin frictional resistance on the pile.

Zc=2.7m

16 kPa

28.2 kPa

28.2 kPa

3.36 kPa

5.92 kPa

5.92 kPa

Overburden pressure

variation closer to the pile

Ultimate skin friction

variation along the pile shaft

155

Solution

i.

The three layers present in the subsurface may be labeled from the top to bottom as L1,

L2 and L3 respectively.

Modification of the friction angle of the LI;

ooo 2933231

From Figure 2.10, Zc/d = 5.5, therefore, the critical depth Zc = 4.4m. the critical depth is

within L1.

Effective overburden pressure at the water table level = 215x = 30 kPa

Effective overburden pressure at the critical depth level = 0.24.430 wsat

= 47.3 kPa.

Below the critical depth level, the effective overburden pressure closer to the pile is

constant with the depth.

Ultimate skin friction: For the bored and cast in-situ concrete piles, the angle of adhesion

a is assumed to be equal to the angle of friction of the soil.

The angle adhesion of L1, a = = 32o

29.0tansin1 a

The angle adhesion of L2, a = = 37o

30.0tansin1 a

wet= 15 kN/m3

Sat= 17 kN/m3

2 m

W. T

7 m

800mm

diameter

bored pile

Medium dense

sand = 32o

Dense sand = 37o

and = 18 kN/m3

Very dense weathered

rock layer = 40o and

= 20 kN/m3

5 m

8 m

156

The angle adhesion of L3, a = = 40o

30.0tansin1 a

Within L1:

At the water table level SF = 3029.0 xv = 8.7 kPa

At the critical depth level SF = 3.4729.0 xv = 13.7 kPa.

SF within L2 = 3.473.0 xv = 14.2 kPa

SF within L3 = 3.473.0 xv = 14.2 kPa

The skin friction distribution can be graphically shown as below:

ii.

The total ultimate skin friction can be estimated by determining the area of the skin friction variation with the depth.

Total skin friction upto the WT = 0.27.82

1x = 8.7 kN

Total SF from 2.0 to 4.4m = 0.24.47.137.82

1 = 26.9 kN

Total SF from 4.4 to 9.0m = 4.40.97.13 x = 63.0 kN

Total SF from L2 = 0.82.14 x = 113.6 kN Total SF from L3 = 0.52.14 x = 71.0 kN

Total SF = 283.2 kN

Zc=2.7m

30 kPa

47.3 kPa

47.3 kPa

8.7 kPa

13.7 kPa

Overburden pressure

variation closer to the pile

Ultimate skin friction

variation along the pile shaft

14.2 kPa

14.2 kPa

14.2 kPa

157

Determination of the end bearing capacity

The pile bottom is pressed against the soil beneath the pile toe and the bearing capacity failure of the soil can occur. However, as the pile toe is at a greater depth below the ground surface, the failure mode is normally the local shear failure and the failure pattern

is similar to the one shown in Figure 2.11.

Figure 2.11– Failure pattern below the pile toe.

As the failure pattern is different from the ones observed for shallow foundations, the same bearing capacity equation, used for the estimation of the ultimate carrying capacity of shallow foundations, may be used with modified bearing capacity factors.

BNNqcNq qcend

2

1

Where, Nc, Nq, N = Bearing capacity factors B = Width of the pile

q = Effective overburden pressure at the toe of the foundation = Unit weight of the material below the pile toe.

It is generally observed that the third term of the above bearing capacity Equation is small compared to other two terms. Therefore, the third term of the above Equation is neglected, if the width of the foundation is not large.

6 – 10 B

2 – 4 B

158

Skempton (1951) suggested the chart given in Figure 2.12 to obtain the bearing capacity factor Nc.

Figure 2.12 – Bearing capacity factor Nc (Skempton, 1951)

It is evident from the chart given above that for a circular or square footing the maximum

value of the bearing capacity factor is 9.0 for L/Br ratio greater than about 4.0. The chart

given in Figure 2.13 is proposed by Berezantzev et al. (1961) for the estimation of the

bearing capacity factor N.

Figure 2.13 – Bearing capacity factor for N (Berezantzev et al., 1961)

The value of the soil should be modified as below before using with the above chart.

159

For driven piles, 2

40'

1

For bored piles, 3'

1

Where '

1 = angle of internal friction prior to installation of pile

Example I

Estimate the ultimate end bearing capacity of the driven 400mm square pile driven 9m

into a clay layer having undrained cohesion 25 kPa.

For normally or slightly over consolidated soils, the undrained capacity is critical,

Therefore, = 0 and Nq = 0.

cend cNq

Since L/B > 4, Nc = 9.0

0.950xqend = 450 kPa

Ultimate end bearing load, Pend = 450x0.4x0.4 = 72 kN.

Example II

Estimate the ultimate end bearing capacity of the 400 mm x 400mm driven pile shown in

the following diagram.

cu = 50 kPa

400 mm x 400 mm driven

pile

9 m

160

Solution.

Equivalent diameter (d) of the pile can be determined by finding the diameter of the pile

having the same cross sectional area as the square pile.

mmxB

d 45142

Modification of the friction angle;

ooo x 3410324

310

4

31

From Figure 2.10, Zc/d = 6, therefore, the critical depth Zc = 2.70m.

Effective overburden pressure at the water table level = 116x = 16 kPa

Effective overburden pressure at the critical depth level = 0.170.216 wsat

= 28.2 kPa.

The modified angle to be used in Figure 2.13 = 2

40'

1

= 36o

N = 90

902.28 xqNq qend = 2538 kPa

Pend = 460 kN.

Exercise

Five boreholes are driven in a proposed building site to investigate the subsurface

condition for a 20-storey building. The subsurface at the site consists of loose silty sand,

stiff clay, completely weathered rock, and fractured rock. A typical subsurface condition

in a borehole and the estimated shear strength parameters are given in the following

wet= 16 kN/m3

Sat= 17 kN/m3

1 m

W. T

8 m

400 mm x 400 mm driven pile

= 32o

161

Figure. As a trial design, 1000 mm diameter bored piles socketed 1m into the fractured

rock is considered. As a design engineer attached to the firm involved in the design,

Estimate the ultimate skin friction of a single pile upto the top surface of the weathered

rock layer.

Empirical correlations

There are large number of empirical correlations that can be used to estimate the skin

friction and end bearing of piles. However, these correlations should be used very

carefully as they are valid under the subsurface condition used to develop them.

Skin Friction

Correlations with the SPT blow counts:

162

Meyerhof (1956, 1976) proposed the following correlation for the estimation of skin

friction (fus):

55Nf mus (kPa)

Where

m = 2.0 for piles with large volumes displacement

= 1.0 for small volume displacement piles

N55 = Statistical average of the blow count in the stratum.

Shioi and Fukui (1982) suggested the following empirical correlations for the estimation

of the skin frictional resistance.

For driven piles: 55,0.2 sus Nf for sand; = 55,10 sN for clay (kPa)

For bored piles: 55,sus Nf for sand; = 55,5 sN for clay (kPa)

Where

Ni,55 = Average blow count in the material indicated for the pile or pile segment length.

Correlations with the Cone Penetration Test (CPT)

Meyerhof (1956) and Thorburn and Mac Vicar (1971) suggested the following

relationship based on the CPT results:

cus qf 005.0 (kPa)

Where

qc = cone penetration resistance in kPa.

When the side friction (qs) of the cone is measured:

sus qf (for small volume displacement piles) and;

sus qtof )0.25.1( (for large volume displacement piles)

End bearing

Correlations with the SPT blow counts:

Meyerhof (1956, 1976) proposed the following relationship for the estimation of the end

bearing capacity.

NB

LNq b

end 38040 (kPa)

Where

N = Statistical average of the SPT N55 numbers in a zone of about 8B above to

163

3B below the pile point.

B = Width or diameter of the pile.

Lb = Pile penetration depth into point-bearing stratum.

Correlations with the Cone Penetration Test (CPT)

Japanese use the following relationship to estimate the end bearing capacity:

cend qq (in units of qc)

Where

qc = Statistical average of the SPT N55 numbers in a zone of about 8B above to

3B below the pile point.

Estimation of the ultimate carrying capacity from the pile driving formulae

This method is commonly used for the estimation of the ultimate carrying capacity of

driven piles. This method is based on two fundamental assumptions:

i. The pile is a rigid body with no elastic deformations; and

ii. The dynamic resistance of the pile is equal to the static resistance of the soil.

Most of the pile driving equations are based on the energy conservation during the

driving process and the equations of motion. Consider the hammer and the pile

immediately before the impact and after the impact shown in Figure 2.14.

Figure 2.14 – Hammer and pile velocities immediately before and after the impact.

u

up

W

Wp vp

v

Immediately before the

impact

Immediately after the

impact

164

The energy transfer from the hammer to the pile and the resulting deformation of the pile

can be diagrammatically shown as given in Figure 2.15.

Figure 2.15 – Energy transfer and the deformation of the pile during a single hammer

blow.

Considering the velocities of the pile and the hammer before and after the impact, and the

deformation of the pile during and after impact following relationships can be obtained.

g

WvWHeE f

2

2

1

The efficiency of impact is

1

2

22

22

22

22

E

E

vgwvgw

ugwugwe

pp

pp

iv

The law of impulse gives :

pp

puv

g

Wuv

g

W

The coefficient of elastic restitution, n, is

p

p

vv

uun

Assuming 0pv , and eliminating u, up, and v,

p

p

ivWW

WnWe

2

The energy left after impact is

p

p

fivfWW

WnWWHeWHeeE

2

2

Various pile driving equations are developed by simplification of the above derived Equation. Some of the commonly used pile driving Equations are given in the following Table 2.1.

165

Table 2.1 Commonly used pile driving equations

Formula

Equation for Ru

Remarks

Senders

S

WH

Engineering

News

CS

WH

C = 1.0 in. for drop hammer

0.1 in. for steam

hammer

0.1 WWp in. for

steam hammer on very

heavy piles

Eytelwein

(Dutch)

PWW

W

S

WH

.

Weisbach

2

2

l

SAE

L

WHAE

L

SAE PPP

Hiley

P

Pf

WW

WnW

CCCS

WHe

2

321

.2/1

See Tables 4.2, 4.3 and 4.4

for values of 321 ,,, CCCe f ,

and n.

Janbu

S

WH

ku

1

dedu CCk 11

WWC Pd 15.075.0 2/ AESWHLe

Danish

21

/2 Pf

f

AEWHLeS

WHe

See Table 4.2 for ef values

Gates

SWHe f 10log6.5 10

Units are inches and tons

(short)

SWHe f 25log0.4 10

Units are metric tons (1000

kg) and centimeters

166

Exercise I Precast concrete piles with 350 mm x 350 mm cross sectional area and a length of 12m

are to be driven for the abutment of a bridge using a 2 ton hammer with a height of drop

of 1m. Estimate the termination set to be achieved if the working load on a pile is 400 kN

and a factor of safety 3 is required against ultimate failure, using Gates method.

Gates equation

SWHeR fult

25log0.4 10

Units are in metric tons (1000 kg) and centimeters. The symbols carry the usual

meanings.

Exercise II

The subsurface at a bridge site consists of a 2m thick recently placed fill followed by

5m thick normally consolidated clay layer, which is underlain by a thick hard

weathered rock layer. 12 m long precast concrete piles with 350 mm x 350 mm cross

sectional area are to be driven at this site for the bridge abutment using a 2 ton

hammer with a height of drop of 1m. Estimate the termination set to be achieved if

the working load on a pile is 400 kN and a factor of safety 3 of is required against

ultimate failure, using

a. Hiley method; and

b. Janbu’s method.

Hiley pile driving formula:

p

pf

ultWW

WnW

CCCS

WHeP

2

321 2/)(

where

AE

LPC ult2

C1 = 3mm C3 = 2.5 mm

n = 0.4 ef = 0.75

W = Drop weight H = Drop height

L = Length of the pile A = Cross sectional area of the pile

Wp=Weight of the pile S = Set of the pile during driving

E = Young’s modulus of the pile material

Janbu’s formula:

S

WH

KP

u

ult

1

167

Where

d

edu

CCK

1

W

WC

p

d 15.075.0

2AES

WHLe

(Assume unit weight of concrete and Young’s modulus of concrete as 24 kN/m3 and 27 x

106 kPa respectively)

168

3.0 Estimation of the settlement of a vertically loaded single pile

As any other foundation, the design of the pile foundations should be safe against

excessive settlements. Therefore, settlement of the pile should be estimated and checked

against the allowable settlement of the foundation. The settlement estimation methods

could be divided mainly into three types:

i. Methods involving empirical correlations;

ii. Semi – elastic approaches involving Load –transfer methods considering the

axial force at various points along the pile shaft;

iii. Methods based on theory of elasticity that involves the use of Midlin (1936)

equations for subsurface loading within the semi-infinite mass; and

iv. Use of the numerical methods such as Finite Element Method

Empirical correlations:

Meyerhof (1959) Method

Based on the field load test results on piles in sandy soils, Meyerhof suggested that the

settlement could be obtained from the Equation [3.1] if the applied load has a factor of

safety more than three against the applied load.

F

db

30 [3.1]

Where

db - diameter of the pile base

F - Factor of safety on ultimate load (Must be > 3.0)

It should be noted here that there is no soil properties nor applied load is in the settlement

estimation equation and hence the validity of this method is highly questionable.

Focht (1976) method

Focht proposed an empirical equation to estimate the settlement of a pile using the

movement ratio, the ratio between the settlement of the pile () and settlement of the pile

acting as a column under the working load. Based on observation of piles in clayey soils

Focht suggested that the use of Equations [3.2] or [3.3] to estimate the settlement of a

single pile.

5.0Col

if col > 8mm [3.2]

and

0.1Col

if col < 8mm [3.3]

169

Methods based on theory of elasticity that involves the use of Mindlin (1936)

equations

Various researchers have used this approach to estimate the settlement of a single pile. In

most of these approaches, the pile is divided into a number of uniformly loaded elements,

and a solution is obtained by imposing compatibility between the displacements of the

pile and the adjacent soil for each element of the pile.

The displacement of the pile are obtained by considering the compressibility of the pile

under the axial loading. The soil displacements are obtained in most cases by using

Mindlin’s equations for the displacements within a soil mass caused by loading within

the mass. The difference between the various methods lies in the assumptions made

regarding the distribution of shear stress along the pile.

The method derived by Poulos and Davis (1968) is described below. The method

assumes a floating or frictional pile in a semi-infinite mass as shown in Figure 3.1.

Figure 3.1 – pile soil model used by Poulos and Davis (1968) for settlement estimation.

It is assumed in almost all the settlement analysis of piles that the pile and soil are stress-

free and that no residual stresses exists in the pile resulting from its installation. This

could be a false assumption for most of the practical situations. However, this error could

be somewhat minimized by selection of appropriate material properties.

If conditions at the pile-soil interface remain elastic and no slip occurs, the movement of

the pile and the soil should be equal. In the solution process only the vertical

displacement compatibility is considered and no lateral displacement of the pile is

considered.

P

D

L

h Soil Young’s

modulus, Es,

and Poisson

ration, vs

Young’s

modulus of

the pile

material is Ep

170

The results of the analysis carried out by Poulos and Davis (1968) are presented in terms

of a parameter referred to as the relative stiffness (k) of pile. The relative stiffness factor

k is defined by Equation [3.4].

s

p

AE

ERk [3.4]

Where RA is the area ratio defined by Equation [3.5]

c

p

AA

AR [3.5]

Where Ap – area of the pile cross section

Ac - Area bounded by the outer circumference of the pile.

Consider a pipe pile of outer diameter of Do and inner diameter of Di as shown in Figure

3.2. The relative are RA is given by Equation [3.6].

2

22

4

4

o

io

A

D

DD

R

[3.6]

Figure 3.2 – Cross section of a pipe pile.

If the pile has a solid cross section without any cavities within it, the area ratio RA is

equal to unity.

Separation of the skin friction and end bearing capacities

The theory presented in Poulos and Davis (1996) can be used to determine the skin

friction distribution along the pile shaft and the hence to separate the skin friction and end

bearing.

A uniform floating pile in a semi-infinite elastic medium, the ratio between the skin

friction and the average skin friction for piles with K=5000 and K=50 are shown in

Figure 3.3. the variation shown in Figure 3 is obtained assuming no-slip condition

between the pile and the soil. It is clear from the Figure 3.3 that the stress distribution

becomes highly non-uniform, when the pile stiffness factor is smaller due larger

settlement of the pile near the top of the pile as a result of high compressibility of the

pile. However, as the pile stiffness becomes higher, the skin friction distribution becomes

more or less uniform. The Poisson ratio of the soil has a negligible effect on the skin

friction distribution.

Do

Di

171

Figure 3.3 – Stress distribution along the pile shaft of a floating pile.

If the elastic modulus of the bearing layer is Eb and the elastic modulus of the material

along the pile shaft is Es, the load transfer curves of end bearing piles, with different

Eb/Es, are shown in Figure 3.4.

Figure 3.4 – Variation of the axial force with the depth of the pile.

Based on the Mindlin (1936), Poulos and Davis (1996) suggested the following

methodology in estimation of the settlement of a single pile.

172

Settlement of a floating pile

According to Poulos and Davis (1996), the settlement of a single pile (ρ) may be

expressed as given in Equation [3.7].

DE

PI

s

[3.7]

Where

P - Applied axial force

I - Settlement influence factor

Es - Elastic modulus of the surrounding material along the pile shaft

D - Diameter of the pile

Settlement influence factor (I)

Settlement influence factor I can be expressed as:

hvko RRRII [3.8]

Where Io – Settlement influence factor for an incompressible pile (k=) in a semi infinite

elastic medium with a Poisson ratio =0.5. Io could be obtained from Figure 3.5.

Figure 3.5 – Settlement influence factor Io

173

Rk, Rv, and Rh are the modification factors, which could be obtained from Figures 3.6, 3.7

and 3.8 respectively.

Figure 3.6 – Modification factor Rk

Figure 3.7 – Modification factor Rv

174

Figure 3.8 – Modification factor Rh

Settlement of an end bearing pile

According to Poulos and Davis (1996), the settlement of a single pile (ρ) may be

expressed as given in Equation [3.9].

DE

PI

s

[3.9]

Where

P - Applied axial force

I - Settlement influence factor

Es - Elastic modulus of the surrounding material along the pile shaft

D - Diameter of the pile

Settlement influence factor (I)

Settlement influence factor I can be expressed as:

bvko RRRII [3.10]

Where Io – Settlement influence factor for an incompressible pile (k=) in a semi infinite

elastic medium with a Poisson ratio =0.5. Io could be obtained from Figure 3.5.

Rk, Rv, and Rb are the modification factors, which could be obtained from Figures 3.6, 3.7

and 3.9 respectively.

175

Figure 3.9 – Modification factor Rb

Estimation of the settlement of piles through layered medium

It is very rarely that the piles are installed through homogeneous medium. In reality, piles

are generally installed through layered soil mediums. Therefore, estimation of the piles

through layered medium should be performed. Figure 3.10 shows the settlement

influence factor (Io) estimated by various methods for a two layer medium with different

moduli ratio. The settlement influence factor (Io) estimated from more sophisticated

methods agree well with that estimated using the weighted average of the elastic moduli

of the two layers. Therefore, weighted average of the elastic moduli of the layered

medium is used in the estimation of the settlement of piles through layered medium as

shown in Figure 3.11 and Equation [3.11]. Similarly, the Poison ratio of the layered

medium is estimated using Equation [3.12].

176

Figure 3.10 – Settlement influence factor (Io) of a two layer medium.

Figure 3.11 – Layered medium

The elastic modulus to be used in the settlement estimation is given by Equation [3.11]

and that for the Poison ratio is given by the Equation [3.12]

hn

h3

h2

h1

Ei

E1 vi

E2 v2

E3 v3

Ei vi

En vn

177

n

i

i

n

i

ii

av

h

hE

E

1

1 [3.11]

n

i

i

n

i

ii

av

h

h

E

1

1

[3.12]

Example I

The thickness and elastic compressibility properties of the soil and rock layers near one

borehole are as follows:

Layer Thickness

(m)

Elastic modulus

(kPa)

Poisson ratio

Organic clay layer 7 2000 0.3

Medium dense sand

layer

8 15000 0.2

Completely

weathered rock

layer

13 50000 0.2

Highly fractured

rock

2 100000 0.2

Bedrock 1 200000 0.1

If a 800 mm diameter bored pile installed near this borehole is socketed 1m into the

bedrock layer, estimate the settlement of the pile under a working load of 2500 kN.

(Elastic modulus of concrete is 31.7 x 106 kPa)

Solution

3174131

1000002135000081500072000

xxxxEavg

22.031

2.022.0132.0873.0

xxxxAvg

100031741

107.31 6

x

E

Ek

Avg

p

178

hvKo RRRII

Io = 0.05

Rk = 1.2

Rv = 0.89

Rb = 0.4

Settlement of a pile is

mmx

xxxx

DE

PI

s

28.031741

)4.089.01.105.0(25001

Exercise

If the drained compressibility parameters, given in Table, are assumed for the subsurface

layers and the bedrock shown in Figure, estimate the expected final settlement of a 600

mm bored pile installed upto the bedrock.

Table

Layer Drained Young’s

Modulus (kPa) Poisson ratio (/) Thickness (m)

Fine sand

layer

10000 0.2 3

NC clay layer 15000 0.3 5

Weathered

rock layer

30000 0.2 6

Bed rock 150,000 0.1

Figure

179

Pile Groups

180

7.0 Design of Pile Groups Subjected to Vertical Compressive Loads

Introduction

Depending on the carrying capacity of individual piles and the working load acting

through the structural elements, such as columns, there are situations that a single pile is

not capable of supporting the structural load. In such situations, it is customary to use a

group of piles to support such structural loads. Like any other type of foundations, the

pile group should also be designed considering:

i. Shear failure of the pile group – should have a reasonable factor of safety

against ultimate shear failure of the soil supporting the group; and

ii. The settlement of the group under the working loads – The settlement of the

pile group under the working load should be less than the allowable settlement

limit of the structure.

General configurations of pile groups are shown in Figure 4.1. When several piles are

clustered as shown in Figure 4.1, it is reasonable to expect that the soil pressures

produced from either side friction or point bearing will overlap as shown in Figure 4.2.

Figure 4.1 – Typical pile group patterns: (a) for isolated pile groups; and (b) for

181

foundation walls.

The resultant pressure intensity will depend on both the pile load and spacing. The

intensity of the stress due to overlapping, will obviously decrease with increased pile

spacing, s. However, it is not practical to have large spacing between piles as the cost of

the pile cap will be high with a larger pile cap. Therefore, center to center spacing (s)

between adjacent piles is a critical parameter in designing pile groups.

Figure 4.2 – Stressed zone surrounding an individual pile and the effect of the group

action.

Consider the stressed zones of an end bearing single pile and a pile group as shown in

Figure 4.3. The width and the depth of the stressed zone beneath a pile group is much

more than that of a single pile in the group. Due to the larger stressed zone beneath the

pile group, the group will undergo a higher settlement than the settlement of individual

piles under the same axial force. Therefore, settlement of a pile group may be critical

even though that of a single pile may be negligible.

182

Figure 4.3 – Extent of the compressed zones beneath: (a) single pile; and (b) Pile group.

Furthermore, due to the large stressed zone beneath the pile group, a weak layer present

may be incorporated in the settlement and ultimate carrying capacity estimation of a pile

group eventhough such layer will not be influencing the behavior of an individual pile as

shown in Figure 4.4.

Figure 4.4 – Effect of a weak soil layer beneath the pile tip: (a) Single pile; and (b) Pile

group.

Ultimate Carrying Capacity of a Pile Group

According to the load transfer mechanism, pile groups can be subdivided into two

categories as shown in Figure 4.5:

i. Free-standing pile group – the pile cap is not in contact with the soil; and

ii. Piled foundation – when the pile cap is in contact with the soil.

Relatively weak soil layer

Strong

soil

layer

Weak

soil

Layer

(a) (b)

Stressed zone

183

Figure 4.5 – Types of pile groups; (a) Piled Foundation; and (b) Free-standing pile group.

In estimation of the ultimate carrying capacity of pile groups, the main concern is the

capacity of the pile group in relation to the summation of the capacities of the individual

piles forming the group. Therefore, the group efficiency (Eg) is defined as given in

Equation [4.1].

n

ii

g

g

Q

QE

1

[4.1]

Where Qg - Ultimate capacity of the pile group

Qi - Ultimate capacity of the ith pile in the group

n - Number of piles in the group.

Pile Groups in Clay

Free-standing pile group

There are mainly two methods of estimation of ultimate capacity of a free-standing pile

group in clay:

i. Using empirical correlations; and

ii. Method proposed by Terzaghi and Peck.

Following Example will be used to discuss the pile group in clayey soils.

Example 4.1

A pile group is formed by driving 9m long 400mmx 400mm precast concrete piles at

1125mm spacing as shown in Figure E4.1. As shown in Figure E4.1, the piles are driven

through a clay layer having cu = 25 kPa.

(a)

(b)

184

Figure E4.1.

It is needed to estimate the ultimate carrying capacity of a single pile in the pile group

given in Example 4.1.

Using the -method,

= 0.95

Therefore, the ultimate skin frictional capacity (pult,skin) of a single pile is given by

Equation [E4.2].

360)944.0(250.1, xxxxp skinult kPa [E4.2]

The end bearing capacity (pult,end) is given by Equation [E7.3].

36259)4.04.0(, xxxp endult kPa [E4.3]

Therefore, the total ultimate capacity of a single pile is 396 kPa.

Equivalent diameter of a single pile is

44.04.0 xx= 451 mm

Estimation of ultimate carrying capacity of a pile group in clay using empirical

correlations

Converse-Labarre formula

One of the empirical correlations very widely used earlier is the one referred to as the

Converse-Labarre formula. However, at present this method is not widely used.

According to the Converse-Labarre method, the efficiency of the pile group is given by

Equation [4.2].

9 m

Cu = 25 kPa 1125 mm

1125 mm

1125 mm 1125 mm

400 mm x 400

mm square

piles

185

mn

nmmnEg

11

901

[4.2]

Where

m - number of rows

n - Number of piles in a row

- Tan-1(d/s) in degrees

d - Pile diameter

s - center to center spacing between piles

Applying the Converse-Labarre formula for the pile group given in Example 4.1,

m = 3

n = 3

=

1125

4511Tan = 21.9o

From the Converse-Labarre method,

Eg = 0.676

From equation [4.1],

9396676.01

xxQEQn

i

igg

= 2409 kN

Feld’s rule

Calculated capacity of a pile is reduced by 16

1 for each adjacent pile, irrespective of the

pile spacing.

Applying the Feld’s rule for the pile group in Example 4.1:

Considering the plan view of the pile group, three types of piles are identifiable, as shown

in Figure E4.4, based on the number of adjacent piles to each type.

Figure E4.4 – Type of piles according to the number of adjacent piles.

Type A piles have 3 adjacent piles;

Type B piles have 5 adjacent piles; and

Type C pile has 8 adjacent piles.

A A

A A

B

B B

B

C

186

Therefore, the group capacity Qg can be estimated from Equation [E4.2].

38616

1814

16

1514

16

131 xxxxxxQg

[E4.2]

25083869722.01 xxxnxQEQ gg kN.

Therefore, from the Feld’s rule, the efficiency of the pile group is 0.722.

The capacity of the pile group predicted by these empirical correlations vary widely and

there is no basis to select one method over the other method as there is no evidence to

support any method.

Terzaghi and Peck Method

This method is the most widely used method to estimate the carrying capacity of pile

groups. This method identifies two ways a pile group can fail: (i) failure of individual

piles in the group; and (ii) failure of the group as a one single block (or commonly

referred to as the block failure mode). Based on the above modes of failure, Terzaghi and

Peck defined the ultimate capacity as the lesser of : (a) The sum of capacities of the

individual piles in the group; or (ii) the bearing capacity of a block encompassing the pile

group as shown in Figure 4.6.

Figure 4.6 – Block failure of a pile group

L

Lg

Bg

Average Cohesion

along the pile

shaft, cavg

Average cohesion of the soil at the base is cb

187

Considering the pile group shown in Figure 4.6, the block failure capacity of the pile

group (Qblock) can be estimated from Equation [4.3].

avgggcbggBlock LcLBNcLBQ 2 [4.3]

Group capacity given in Example E4.1 can be obtained using the Terzaghi and Peck

method.

Bg = 1125x2 + (400x2)/2 = 2650 mm

Lg = 2650 mm

L = 9 m

cb = cavg = 25 kPa

25965.265.229.82565.265.2 xxxxQBlock 3947.5 kN.

93969

1

xQi

i 3564 kN

Therefore, the group capacity is 3564 kN and the failure mode is the individual pile

capacity mode.

Terzaghi and Peck method assumes that the failure takes place either due to individual

pile failure or due to block failure mode. Experimental results have shown the existence

of these two types of failure modes. For closer pile spacing, block failure becomes critical

whereas for wider spacing between piles in the group, individual pile failure becomes

critical. Terzaghi and Peck method assumes that there is a certain pile spacing below

which block failure takes place and for pile spacing higher than the critical spacing,

individual pile failure takes place. However, there is no experimental evidence, to support

such an abrupt transition from block failure to individual pile failure mode. In order to

obtain a more realistic estimate of the ultimate load capacity a group, the empirical

relationship given in Equation [4.4] could be used.

22

1

22

111

Blockg QQnQ [4.4]

Qg = Ultimate capacity of the group

Q1 = Ultimate capacity of a single pile

QBlock = Ultimate capacity of the block failure

n = Number of piles in the group

Pile Groups in Layered Medium

If the piles in the group are installed through layered soil medium, the portion of the piles

through the clay layers should be checked against block and individual failure. Individual

capacities of the piles should be considered for the portion of the piles through the sandy

soil layers. Consider the pile group through a layered soil medium as shown in Figure

188

4.7.

Figure 4.7 – Pile group through layered soil medium.

Bearing capacity factor for soil failure at the bottom of the block.

Pile Groups in Sandy Soils

There is ample evidence to believe that the ultimate carrying capacity of pile groups in

sandy soils is more than the summation of the ultimate capacities of individual piles in

the group. This is mainly due to the compaction of the sandy soils from the vibration

created by vibration generated due to pile driving. Figures 4.8 and 4.9 show the variation

of the group efficiency factor of pile groups in sand with the space-to-diameter ratio of

the pile group.

Sandy soil

(Skin friction of individual piles

through sand layer should be

considered)

Clayey soil

(Block failure mode should also be

considered through the clay layer)

Sandy soil

(Skin friction of individual piles

through sand layer should be

considered together with the

individual point bearing capacity)

189

Figure 4.8 – Variation of the group efficiency factor with the space-to-diameter ratio of

the piles in sandy soils.

.

190

Figure 4.9 – Variation of the group efficiency factor with the space-to-diameter ratio of

the piles in sandy soils (Vesic, 1969)

The broad conclusion drawn from the above data is that unless sand is very dense or the

piles are widely spaced, the overall efficiency is likely to be greater than 1. The

maximum efficiency is reached at a spacing of 2 to 3 diameters and generally ranges

between 1.3 and 2.

191

Settlement of Pile Groups

As shown in Figure 4.3, the influence zone below the pile bottom is very high compared

to that of a single pile. As a result the volume of soil compressed is more and hence, the

settlement of a pile group can be appreciable compared to a single pile even if the load

acing on a single pile in both cases is of the same order. Therefore, settlement analyses of

pile groups become important. Settlement estimation methods of pile groups can be

divided into following categories:

i. Empirical correlations;

ii. Equivalent raft method;

iii. Interaction between individual piles; and

iv. Finite Element Method.

Empirical Correlations

There are certain empirical correlations, developed by various researches based on their

experience, to estimate the settlement of pile group.

Skempton (1953) suggested the empirical relationship given in Equation [4.5] to estimate

the settlement of a pile group.

2

2

1 12

94

B

BG

[4.5]

Where

G - Settlement of the pile group

1 - Settlement of a single pile under the average load

B - Width of the pile group in feet

For driven piles in sand, Meyerhof (1959) suggested the relationship given in Equation

[4.6] to estimate the settlement of pile groups.

2

1 11

35

r

ss

G

[4.6]

However, there is very little experimental data to support these empirical methods and

these methods are not very often used at present.

Equivalent Raft Method

The principle behind this method is that the settlement of a pile group is equivalent to that

of a raft placed at a certain level in the subsurface. The level, at which the raft is placed,

is decided based on the subsurface condition and the load carrying mechanism of the

piles in the group. Some suggested levels for placing the raft are shown in Figure 4.10.

192

(a)

(b)

(c)

Figure 4.10 – Placement of an equivalent raft for settlement estimation: (a) friction piles

in soft soil; (b) piles through a soft layer end bearing on a hard granular material; and (c)

Pile group end bearing on bedrock.

Once the equivalent raft is placed and the load acting on the pile group is applied on the

raft, the problem becomes that of a shallow foundation. The settlement of the equivalent

raft may consist of immediate settlement and consolidation settlement components. If the

Equivalent raft

Soft clay

Bedrock

Equivalent raft

Distribution

of load at 1:4

L 2L/3

Soft clay

Dense granular soil

L

Equivalent raft

Distribution

of load at 1:4 2L/3

Soft clay

193

pile group is in clayey soils, consolidation settlement should be estimated in addtion to

the immediate settlement of the raft.

Estimation of the Immediate Settlement of the Group

Immediate settlement (i) at the center of the equivalent raft could be estimated from

Equation [4.9].

Fp

u

ni IIE

Bq

212

[4.7]

Where

21

1

21FFI p

[4.8]

qn - Net pressure from the equivalent raft

B - Width of the equivalent raft

Eu - Undrained Young’s modulus of the clayey soil

- Undrained Poisson ratio of the clayey soil taken as 0.5

F1 & F2 - Factors from Figure 4.11

IF - Fox’s correction factor (Figure 4.12)

H - Thickness of the compressible material below the equivalent raft.

Figure 4.11 – Factors F1 and F2

H/B

194

Figure 4.12 – Fox’s depth correction factor IF.

It is extremely important to obtain a reliable value for the Young’s modulus of the soil

under the equivalent raft. Due to sample disturbance, unrealistically low Young’s

modulus values are obtained from stress-strain curves obtained from conventional

unconfined or triaxial compression tests in the laboratory. In-situ tests, such as plate load

tests done in boreholes or trial pits, may be used to obtain Young’s modulus, which

represent the field conditions much better. The stress-strain behavior of soil is highly

nonlinear. Therefore, the usual practice is to draw a secant AC to the stress-strain curve

corresponding to a compressive stress equal to the net foundation pressure at the base of

the equivalent raft, as shown in Figure 4.13.

Figure 4.13 – Estimation of the secant modulus of soil under the equivalent raft.

If the Young’s modulus of the soil under the equivalent raft is not constant with the

depth, Equation [4.9] cannot be used.

195

Development of negative skin friction

The load carrying capacity of piles come from skin friction and end bearing. The

development of the skin friction and the associated deformation of the soil and the pile

could be simply illustrated as shown in Figure 5.1 using a typical pile element with the

corresponding soil elements in contact with it. Figure 1(a) shows the equilibrium position

of the pile and the corresponding soil elements before loading the pile. Figure 5.1(b)

shows the location of the pile and soil elements after application of an axial force of P1.

(a) (b) (c)

Figure 5.1. Development of ultimate skin friction: (a) pile at rest condition with one

pile element in contact with the adjacent soil element; (ii) During loading

of the pile; and (c) slipping of the pile and soil elements after development

of the ultimate skin friction.

Figure 5.1(c) shows the location same elements after slipping between the pile and the

soil. When the slipping between the pile element and the soil element occurs, the

frictional resistance acting on the pile has reached the ultimate value (skin, ult). The shear

deformation required to mobilize the ultimate skin friction is considered to be relatively

small. Since the direction of the skin friction is opposite to the direction of the applied

axial load, the frictional resistance is referred as skin frictional resistance (or positive

friction). The skin friction is generated due to the relative deformation between the pile

and the soil elements.

If the soil surrounding the pile moves in the downwards direction relative to the pile, the

direction of the frictional force acting on the pile is in the downward direction. As a

result, the direction of the frictional force is same as that of the applied load. Therefore,

such frictional force acts as a load on the pile. This phenomenon is referred to as the

negative skin friction.

The surrounding soil can move in the downward direction as a result of the consolidation

settlement of the surrounding soil due to: (i) lowering of the ground water table; (ii)

surcharging of a clay layer as a result of placing a fill layer; and (iii) increased pore water

pressure in the vicinity of a driven pile. A portion of the pile must be fixed against the

vertical movement for the development of a significant negative skin friction on the pile.

If the pile moves with the consolidating soil, there is no negative skin friction developed.

In Sri Lanka, bored piles are end bearing and driven piles are driven to a strong bearing

Pile

element

skin skin,ult

Pile

Soil elements

in contact

with the pile

element

P1 P2

196

layer. Therefore, significant negative skin friction can be developed. Most piling sites in

Sri Lanka with a soft surface soil layer are filled prior or after installation of the piles.

Such soil deposits, which are normally consolidated undergoes large amount of

consolidation settlement over a period of time after placing such a fill. The time taken for

the completion of the major portion of the settlement depends on the thickness and the

consolidation properties of the compressible soft soil layer. Another significant feature of

the ground condition in Sri Lanka is a presence of a relatively residual soil formation

generally consisting of thick weathered rock layer overlying the bedrock. Therefore, a

typical subsurface condition with a possibility of development of negative skin friction

consists of a fill layer underlain by a normally consolidated soft soil layer followed by a

residual weathered rock layer on the bedrock. A typical subsurface is shown in Figure

5.2.

Figure 5.2. Typical subsurface profile and development of negative and positive skin

friction on end bearing piles

The soil upto the bottom of the consolidating soft soil layer is moving in the downward

direction. However, the amount of downward movement is maximum at the ground

surface and zero at the bottom of the soft soil layer. If it is assumed that the pile is

restrained from moving at the bottom, the downward movement of the surrounding soil

generates a drag force at the top of the pile upto a certain depth where the relative

movement of the surrounding soil with respect to the pile is in the downward direction.

Beyond that depth, the movement of the surrounding soil is in the upward direction with

respect to the pile and, therefore, positive friction is developed on the pile. The plane

separating the positive and negative skin friction is referred to as the ‘neutral plane’.

However, if the pile is free to move at the bottom, the neutral point is also moving in the

downward direction and the neutral point is established at the location where the relative

downward movement of the surrounding soil with respect to the pile is zero. In such

situations, the drag force becomes a dragdown force and the pile settles under the

negative skin friction.

Design of piles against negative skin friction

It is very essential that the designer should investigate whether there is a possibility of

Residual soil

layer

Fill layer

Consolidating

soil l layer

Bedrock

Ground surface immediately

after placing the fill

Layer boundaries after

consolidation

Neutral plane

197

development of negative skin friction at the site concern. Presence of a soft layer or a

clay layer alone is not a reason for development of negative skin friction. Possibility of

placing a fill within the site area and lowering of the ground water table should be given

due consideration. If a cause for consolidation of the soft layer is not present or the soft

soil layer is sufficiently overconsolidated to prevent significant consolidation settlement

then the negative skin friction should not be considered in the design. It is evident from

the above discussion that an effective stress method such as -method should be used in

estimation of the negative skin friction acting on a pile. It was also discussed that the

negative skin friction acts only upto the neutral plane from the pile top and beyond that

positive friction is developed. However, it is very difficult to locate the neutral plane for a

multi-layer soil subsurface commonly encountered in Sri Lanka. Therefore, it is advisable

to estimate the negative skin friction upto the bottom of the consolidating soft layer using

an effective stress method such as -method, which gives the negative skin friction

(skin,neg):

/

, vnegskin

Where /v is the effective overburden pressure at the

Meyerhof (1976) proposed the skin friction factor () for driven piles given in Figure 5.3.

It should be noted here that the value obtained from Figures 5.3 agrees well with the

experimental case studies present in this paper.

Figure 5.3 Negative skin friction factor for piles driven into soft to firm clays,

Meyerhop (1976)

It should be noted here that considering the drag load as a load acting on the pile for

determination of the factor of safety could be conservative. Therefore, it is suggested that

the required ultimate carrying capacity of the pile be obtained by multiplying the working

198

load by a normal factor of safety. Once the carrying capacity of the pile is obtained,

neglecting the skin friction from the consolidating layer and other layers above it, it

should be divided by the working load plus the drag force to check whether a reasonable

factor of safety exists.

199

Pile Load Test

200

8.0 Pile Load Test

Testing of piles is done mainly due to the following reasons:

i. To evaluate the performance of pile at the preliminary or later stages in terms

of settlement and carrying capacity;

ii. To assess the structural integrity of the pile; and

iii. To obtain additional information required for the pile design such as: total

skin frictional capacity; distribution of the skin friction along the pile shaft

and mobilized end bearing.

Testing of the piles could be carried out at mainly two different stages:

i. Testing of test piles prior to the construction of working piles; and

ii. Testing of piles during construction stage of the working piles.

Piles tested during the construction stage could be further subdivided into (a) Piles for

preliminary testing and (b) Routine proof testing of piles.

The first type of testing is mainly conducted on test piles solely constructed to assess the

performance of the piles to get some inputs to the design and construction of the pile

foundation. The inputs that are generally obtained by testing test piles, includes design

strength parameters, compressibility of the pile and the suitability of a particular design

method. Since the construction and testing of the test pile(s) is very costly, the decision to

construct and test the test pile(s) should be done considering

the importance and magnitude of the foundation,

complexity of the subsurface condition at the site,

possible influence of the installation process on the surrounding area, and

the anticipated problems during the construction stage.

Number of test piles in a particular site may be decided mainly on the variability of the

subsurface condition within the site. The information obtained by testing the test pile(s)

may give the foundation designer the necessary confidence to go for an economical

design and construction methodology giving considerable savings to the client.

Moreover, the information gathered during the construction of test pile(s) could be used

very effectively to plan the construction process and the quality controlling program to be

adopted during the construction of the actual pile foundation. However, the results of the

test pile(s) should be considered together with the possible variability of the subsurface

and the bedrock condition within the site.

The second type of testing is mainly done during construction stage and they could be

further subdivided into

(a) Piles for preliminary testing and

(b) Routine proof testing of piles.

Preliminary test piles are constructed at the beginning of the construction stage to serve

as the test piles. The information intended to gather from a preliminary test pile are same

as that of a test pile. Since the preliminary test piles are constructed using the same

machinery and the methodology, they represent the working piles better than the test piles

constructed and tested prior to the construction of the working piles. However,

continuous construction of working piles after construction of the preliminary test piles

201

could be risky if the preliminary test piles prove that the design assumptions are not

correct. This happens as the cast in-situ bored piles cannot be tested until concrete gains a

reasonable strength, preferably until 28 days after casting, working piles constructed

during the time period between casting the test piles and testing them may not meet the

design requirements. Therefore, such piles should be supplemented with other piles or

replaced with new piles. The author has observed few such instances and found that the

cost of remedial work required is very high,

The impotence of following a well planned quality control program during construction

of bored piles was explained earlier. However, the quality of the constructed pile

foundation should be verified by subjecting selected piles to load tests. The number of

piles to be tested depends on many factors such as:

importance and magnitude of the foundation,

variability of the ground condition across the site,

results of the quality control program, and

factor of safety used in the deign.

Moreover, based on the other post construction testing such as, integrity testing of bored

piles, piles for load testing may be selected. Testing of bored and cast in-situ piles may be

broadly classified into: (i) load testing; and integrity testing.

Load Testing Piles

In load testing of piles, the pile is loaded upto a proof load or upto failure. Through load

testing it is aimed at establishing the load deflection behavior of the pile upto a proof load

or ultimate carrying capacity of the pile under the applied load. Depending on the type of

load applied on the pile, the load testing may be divided into following categories:

i. Compression load testing

ii. Tension load testing; and

iii. Lateral load testing.

Most of the bored piles constructed in Sri Lanka are subjected to compressive loads.

Therefore, most of the load testing carried out are compression load tests and only

compression load tests are considered in this lecture. However, there are situations, where

lateral load tests are carried out on bored piles to investigate the lateral load carrying

capacity, especially the piles, which are subjected to lateral earth pressure.

Compression load testing can be classified into three major types depending on the rate of

loading:

i. Static Load Testing;

ii. Dynamic load testing; and

iii. Statnamic load testing.

Static load testing can be further divided into two types: Conventional static load testing;

and static load testing using Osterburg cell.

202

Testing of bored piles is very often carried out to investigate the load-settlement behavior

of the tested pile upto a proof load. Normally in most cases, the pile is loaded upto about

1.5 times the working load to establish the load-settlement behavior. The termination of

the load testing at 1.5 times the working load is mainly due to the cost associated with

continuation of the loading beyond that level. If the pile doesn’t reach the ultimate failure

before the final test load, the ultimate carrying capacity of the pile cannot be obtained

directly from static load testing. Therefore, it is not possible to obtain the Factor of Safety

(FoS) of the pile against the shear failure of the soil surrounding the pile. Criteria for

satisfactory performance of the pile are specified based on the settlement of the pile at the

working load and the proof load. The specified settlement at the working load and the

proof load varies from testing standard used.

Advantages and Disadvantages of Different Methods of Load Testing

The information that can be gathered from load testing and the accuracy of the

information obtained vary from the type of the test used. More importantly, the cost of

testing depends very much on the method of the test adopted. Therefore, the selection of

the type of load test should be done based on the cost, information needed from testing

and the required accuracy of the information.

Conventional static load test produces the load – settlement curve as the final result of the

test. Normally the settlement at a given load and the likely plastic settlement

corresponding to an applied load could also be estimated, if unloading of the pile is also

done during testing. Separation of the total load into the skin friction and end bearing is

not directly available from a static load test result. Eventhough there are various methods

of separation of end bearing and skin friction such as method proposed by Van Weel

(1957) and Chin (1970), the accuracy of the predictions vary widely. If accurate skin

friction and end bearing information are needed, instrumentation of the pile with strain

gauges at different levels should be done. However, this is feasible only in the case of test

piles as the instrumentation of the pile should be done during installation. The load –

settlement curve obtained from static load test may not indicate the structural integrity of

the pile, if structural failure of the pile doesn’t take place at defective sections. Various

researchers have shown that the shape of the load-settlement curve can be interpreted to

get more information related to the tested pile. Thilakasiri (2007) attempted to describe

the reasons for commonly observed variations of the shape of the load-settlement curves

obtained by testing bored piles by using the Elastic Shortening Line (ESL). More detailed

discussion on the interpretation of the load – settlement curves obtained from static load

testing is presented in Chapter 6.

Static load test using Osterburg cell has the main advantage of obtaining the mobilized

skin friction and end bearing of the pile separately. Furthermore, load deflection curves

for the shaft and the end can be separately obtained. Since the Osterburg test doesn’t

require external reaction system, it can be performed under low headroom conditions.

However, the maximum load the pile can be tested is twice the minimum of ultimate skin

friction or the ultimate end bearing.

In contrast to a traditional load test, dynamic load test using the Pile Driving Analyzer

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(PDA) yields direct information regarding the skin friction distribution along the pile

shaft, mobilized end bearing capacity, structural integrity of the pile etc. in addition to the

static load - settlement curve of the pile. Since the static behavior of the pile is estimated

from dynamic loading, the accuracy of the results of the dynamic load test depends on

number of factors such as the accuracy of the data gathered, experience of the person

analyzing the PDA data, data analyzing model used etc. Therefore, it is a common

practice to load test a certain number of piles using both static and dynamic methods and

calibrate the parameters used in interpretation of the PDA data using the load – settlement

curve obtained from the static load test.

Following facts should be considered in comparing the load – settlement curves obtained

from static and dynamic load tests.

i. Certain amount of creep settlement is included in the static load settlement

curve as the load is kept constant for a certain specified time period whereas

during dynamic load testing immediate response of the pile is obtained.

ii. Settlement measurement during static load testing is done using three to four

dial gauges and the average is taken as the settlement, which may be

erroneous if the difference between the dial gauge readings is high.

Various correlation studies were done throughout the world to investigate the accuracy of

the dynamic load tests. Most of the correlation studies were done on driven piles as the

PDA was originally developed to test driven piles. Most of the studies to compare static

and dynamic test results are done only considering the failure load predicted by the two

methods. Likins and Raushce (2002) showed that the Davisson failure criterion is the best

method to use in estimation of the failure load in the comparison studies.

Factors Affecting the Selection of the Load Testing Method for a given Site

Following factors should be considered in selecting the type of test to be used for load

testing of piles in a particular site.

i. The magnitude of the test load or the expected ultimate carrying capacity of

the pile – Depending on the magnitude of the test load the reaction load in the

static load test, input dynamic energy in the dynamic load test and the weight

of the cache system in the statnamic load test may vary. Economy of

providing the required loading method should be given due consideration in

selecting the type of the load test. Static load testing using the Osterburg Cell

test depends only on the relative magnitude of the ultimate skin friction or end

bearing, as the maximum load that can be applied on the pile during Osterburg

Cell test is twice the minimum of ultimate skin friction or ultimate end

bearing.

ii. The type of information that is expected to be gathered from load testing and

the level of accuracy of the required information – Certain test piles demands

more information from it than just the load – settlement curve. For example,

skin friction distribution along the pile shaft, mobilized end bearing, soil

stiffness of the surrounding soil, and integrity of the pile can only be obtained

through a dynamic load test. If more accurate such information is required,

use of static load test on an instrumented pile may be preferred.

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iii. The environment within the site and surrounding it – Use of a static load test

to test piles in water, for example offshore piles, may be extremely difficult.

In such situations the use of the dynamic or statnamic load test may be

preferred. On the other hand if the space available is not sufficient to construct

a loading platform or reaction piles, performance of a static load test may not

be possible. On the other hand, if the ground vibration in a build up area is a

concern, dynamic load testing may not be possible.

iv. Relative cost involved with different types of load testing methods – There is a

significant difference between the costs required to conduct different type of

test. Moreover, preplanning is essential in conducting instrumented pile load

tests such as Osterburg cell test and static load test on instrumented piles.

Static Load Test

Static load testing is the conventional and most reliable method of load testing of piles. In

practice, load testing on piles are performed to either prove that the pile can safely bear a

design load or to establish a design load based on the ultimate pile bearing capacity

obtained from the test. In this test, load is applied at the top of the pile while monitoring

the settlement of the pile top. Based on the way the load is applied at the pile top, two

types of testing methods are followed, namely,

Constant Rate of Penetration (CRP) test and

Maintained Load (ML) test.

In the CRP test, compressive force on the pile is progressively increased to cause the pile

to penetrate the soil at a constant rate until failure occurs. According to BS 8004, the

penetration rate for friction piles in clay is 0.75mm/min whereas that for end bearing

piles in granular soil is 1.55mm/min. However, AS 2159 (1996) specifies penetration rate

for friction piles in clay as 0.5mm/min whereas that for piles in granular soil as 2.0

mm/min. AS 2159 states that the above rates of penetration could be halved or doubled

without significantly affecting the results of CRP tests. However, it should be emphasized

that the load test should be performed adhering to a certain testing specification and

commonly used testing specifications in Sri Lanka include BS 8004, ASTM D1143, and

AS 2159 etc. The testing specification to follow should be agreed upon when the

contractual agreement is reached.

The CRP method is essentially a test to determine the ultimate load on a pile and is

therefore, applied only to preliminary test piles or research type investigations. In the ML

test, load is increased in steps to some multiple, for example 1.5 times or twice the

working load, with the time-settlement curve recorded at each stage of loading and

unloading. At each loading step, the load is maintained constant until the rate of

settlement of the pile is smaller than a specific value, for example 0.25mm/hr. As it is

evident from the testing procedure, the CRP test could be finished within a short time

period than the ML test. Therefore, sometimes when the load-settlement behavior and the

ultimate carrying capacity are both needed, the pile could be tested using ML upto the

test load and thereafter continue loading to failure at a constant rate of penetration. A

study of the load settlement curves of both the ML and CRP methods show that the

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ultimate load portion of each test agrees reasonably well. However, there are differences

prior to the ultimate region. This may be attributed to the ML test including greater soil

creep effect than the CRP test because of its longer duration.

As described previously, the test load applied on the pile should be decided based on the

factors such as possible negative friction and variation in the cutoff levels between the

test piles and working piles.

Conventional Static Load Testing of Bored Piles

Setting up of the Test

The static load is applied on the pile by jacking against a reaction system. The reaction

system is referred to as an arrangement of kentledge, piles, anchors or rafts that provides

the resistance against which load on the pile is applied. Kentledge and reaction piles are

the most widely used reaction systems. Pictures of two kentledges reaction systems are

shown in Figures 6.1(a) and 6.1(b) while that of a reaction pile system is shown in Figure

6.2. It may consist of precast concrete blocks, precast concrete piles, cast-iron blocks, or

any heavy dead weight. The kentledge is supported on a platform or grillage. The

platform or grillage is supported on the ground or on specially prepared foundations,

which are well clear of the test pile. Testing specifications put the limit on the minimum

distance between the test pile and the closest point of the foundation of the kentledge.

The foundation, supporting the kentledge, should be placed on stable ground and should

not give rise to any differential settlement. Figure 6.3(a) shows a situation where

kentledge had collapsed during a static load testing of a pile due to bearing failure of the

soil supporting the kentledge. The center of gravity of the kentledge is on the axis of the

test pile and the hydaulic jack should apply the load coaxial with the pile. Since the

potential energy stored in the kentledge under a large axial force is high, the kentledge

should be stable to avoid failure.

Figure 6.1 - Kentledge reaction systems used for static load testing of piles.

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Figure 6.2 - Reaction system consisting of anchor piles and beam systems

(a) (b)

Figure 6.3 - Failure of Reaction systems (a) Failure of a kentledge; and (b) Failure of

tension bar system in reaction pile (Courtesy of Federation of Piling Specialists)

The total effective weight of the reaction system providing resistance to the hydraulic

jack should be at least 1.2 times the maximum test load intended to be applied on the pile.

If the weight of the reaction system is estimated from the volume and density of the

constituent materials, an adequate factor of safety against any variations should be

allowed. Failure of the reaction system could be disastrous and two occasions of failures

of kentledge reaction system and reaction pile systems are shown in figures 6.3(a) and

3.3(b) respectively.

The top of the pile should be cleaned to remove the weak contaminated concrete upto a

suitable depth below the cutoff level. If the elevation of the top level of the cleaned pile

head is low, it should be built up to the necessary height using a suitable grade of

concrete. Built up length of the pile should be of necessary strength to resist the axial

load at the pile top without undergoing excessive deformation. The built up section of the

pile should be coaxial with the original pile and the top surface of the pile should be

perpendicular to the axis of the pile. The arrangement shall be such that none of the test

load is carried by the ground, under the cap. The pile top should have enough area to

place the hydraulic jack and other deflection monitoring equipment. The top surface of

Anchor pile

Tested pile

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the prepared pile head should be smooth and be free from any projections that might

cause stress concentration and crushing of concrete at the contact points.

The settlement monitoring devices such as mechanical dial gauges, and/or LVDT’s

should be supported on a reference beam, which is not affected by the deformation of the

soil during application of the load on the test pile. As the soil surrounding the pile shaft is

subjected to deformation during application of the axial force on the pile, the ground

supports of the reference beam must be placed away from the loaded pile, and the

foundations of the reaction system. The axis of the settlement monitoring devices should

be vertical to measure the vertical deflection of the pile and should possess an accuracy

specified by the testing specification. It is suggested that four dial gauges (or LVDTs)

must be placed on the pile head to monitor the settlement during load application and

placement of the dial gauges should be selected to catch differential settlement of the pile

head. The displacement measuring devise should have a sufficient travel to the maximum

expected differential settlement between the pile and the reference beam and should

possess a very low sensitivity against the possible environmental changes. The

deformations of the reference beam should be cross checked with respect to a nearby

fixed datum using suitably accurate survey equipment such as precise level. The entire

settlement monitoring system should not be subjected to any external deformations due to

nearby construction activities, vandalism or natural causes such as sunlight, wind etc.

In place of the reference beam and the displacement measuring devises, a wire suitably

tensioned between two stakes could be used to measure the deformation of the test pile.

The settlement is measured by the relative deflection between the wire and a graduated

scale fixed onto the pile.

The load is applied on the pile through a hydraulic jack placed between the pile head and

the reaction system. The jack used should have the required capacity to provide the test

load on the pile. The plunger of the jack should have enough run to undergo extension

equal to the downward deformation of the pile head and the upward deflection of the

reaction system. The total capacity of the jacks used to load the pile should exceed 20%

or more of the required maximum test load.

The load measuring devise may consist of a proving ring or load cell in addition to the

pressure gauge of the hydraulic jack. The load cell or the proving ring used to measure

the settlement of the pile head should have been calibrated recently and the certificate of

calibration performed by an approval agency must be produced. The readings of both the

load measuring device and the pressure gauge should be noted but only the reading of the

load measuring device should be used in interpretation of the measured load. The

pressure gauge readings are required for error checking of the measuring system. A

schematic diagram showing a typical pile load system is shown in Figure 6.4.

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Figure 6.4 - Schematic diagram showing arrangement of the loading system and

measuring devises in a typical static load test

INTRPRETATION OF LOAD-SETTLEMENT BEHAVIOR OF PILES

The main final result obtained from the static load testing of piles is the load –settlement

behavior of the tested pile. The load – settlement curve thus obtained should be used to

estimate the carrying capacity of the pile. In estimation of the carrying capacity of a pile,

the possible failure of the pile due to serviceability failure of the pile, ultimate

geotechnical failure of the pile, or structural failure of the pile should be considered.

Serviceability failure is generally defined with respect to a certain threshold settlement

under loading and unloading of the pile. The threshold of the settlement specified by

varies depending on the standard considered and the type of testing method. AS 2159

(1995) specifies the acceptance criteria of the static compression load test as given in

Table 6.1

Table 6.1 Compression load test acceptance criteria (AS 2159 (1995))

Load Maximum deflection (mm)

Serviceability load 15*

and (after removing serviceability load ) 7*

1.5 x design action effect 50

And (after removing 1.5 x design action effects 30 * Movement to include no more than 3mm creep over 5h (after load has been in place for

15 min.)

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AS 2159 (1995) specifies the acceptance criteria given below when the pile is tested

using dynamic methods:

The pile head movement does not exceed:

(i) 12 mm at design serviceability load; and

(ii) 35 mm when the load is 1.5 x design action effect.

Design action effects mentioned in above may include other possible loading on the pile ,

such as: dragdown forces and additional force due to earthquake etc., during its working

life.

ICTAD/DEV/16 (1997) gives its performance specification for maintained pile load test

as given below:

For load cycle upto 1.0 x Working Load:

Maximum allowable gross settlement = 12 mm

Maximum allowable net settlement = 6 mm

For load cycle upto 1.5 x Working Load:

Maximum allowable gross settlement = 25 mm

Maximum allowable net settlement = 12 mm

The above serviceability criteria may vary depending on the type of structure supported

by the piles. For example, certain structures may tolerate much larger settlements than

what is specified in the above criteria. Even though such performance criteria of piles are

important for design purposes, they do not provide any indication of the capacity of the

pile.

The true ultimate failure of the pile is defined as the load corresponding to point in the

load – settlement curve, where settlement continues to increase without additional

increase in the load (point A in Figure 6.5). This condition may be very difficult to

achieve especially with end bearing bored piles socketed into the bedrock. There are

other definitions of ultimate capacity of piles such as:

1. The load beyond which there is an increase in gross settlement

disproportionate to the increase in load (point B in Figure 6.5).

2. The load indicated by the intersection of tangent lines drawn through the

initial, flatter portion of the gross settlement curve and the steeper portion of

the same curve (point C in Figure 6.5).

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Figure 6.5 Typical load – settlement curve showing ultimate load based on some failure

criteria.

The location of points A and B depend on the personal judgment of the person

interpreting the load – settlement curve. The drawing of the initial tangent and the tangent

of the flatter portion of the curve also depends on the personal judgment. Moreover, the

scale of the graph might have a certain influence on the ultimate capacities determined by

these methods. A good method for estimation of the ultimate capacity should be

independent of the scale effects and the personal judgment of the interpreter.