evaluation of failure modes for concrete...

DEGREE PROJECT, IN , SECOND LEVELCONCRETE STRUCTURES

STOCKHOLM, SWEDEN 2015

Evaluation of Failure Modes forConcrete Dams

LISA BROBERG & MALIN THORWID

KTH ROYAL INSTITUTE OF TECHNOLOGY

SCHOOL OF ARCHITECTURE AND THE BUILT ENVIRONMENT

Evaluation of Failure Modes for ConcreteDams

Lisa Broberg & Malin Thorwid

June 2015TRITA-BKN. Master Thesis 455, 2015ISSN 1103-4297,ISRN KTH/BKN/EX–455–SE

c©Lisa Broberg & Malin Thorwid 2015Royal Institute of Technology (KTH)Department of Civil and Architectural EngineeringDivision of Concrete StructuresStockholm, Sweden, 2015

Abstract

The safety of a concrete dam is ensured by designing according to failure criteria, forall combinations of loads using safety factors. Today in Sweden, RIDAS, the Swedishpower companies’ guidelines for dam safety, is used for the design of dams and isbased on BKR, the National Board of Housing, Building and Planning. Swedishdams are designed to resist two global failure modes; sliding and overturning. Com-bination of failure modes, that should be considered in the design of concrete dams,is however fairly unknown. Since 2009 the Eurocodes was adopted and came intoforce 2011. The Eurocodes have replaced BKR in the design of most structures inSweden where the partial factor method is used to ensure safety in the design.

The objective of this report was to examine if the design criteria for concrete damsin today’s condition are enough to describe real failure modes. The other objectivewas to analyse if Eurocode is comparable to RIDAS in dam design. The statedquestions were answered by performing a literature study of known dam failuresand analytical calculations for different types of concrete gravity dams, with varyinggeometry and loading conditions. The programs CADAM and BRIGADE were alsoused as calculation tools to further analyse if failure occurred as expected.

The results from the analytical calculations together with the performed FE anal-ysis indicate that limit turning does occur and often generate lower safety factorscompared to overturning. Limit turning is similar to overturning failure although itaccounts for material failure in the rock. This design criterion is therefore, highlydependent on the quality of the rock and requires investigations of the foundationto be a good estimation of the real behaviour of the dam body.

From the compilation of reported failures the conclusion was that the current de-sign criteria are adequate. However, the real challenge lies in ensuring that theconstruction of dams is correctly performed to fulfil the stated criteria. A transitionto Eurocode appears to be reasonable for the stability criterion. A modification ofthe partial factors is nevertheless necessary to obtain result corresponding toRIDAS, especially for the overturning criteria.

Keywords: Gravity dams, concrete, design criteria, RIDAS, Eurocode, limit turn-ing

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Sammanfattning

För att uppnå säkra dammkonstruktioner, för alla lastkombinationer, dimensionerasdammar enligt bestämda brottvillkor som ska uppfylla en viss säkerhetsfaktor. Idaganvänds RIDAS, för dimensionering av dammar i Sverige. RIDAS Kraftföretagensriktlinjer för dammsäkerhet, är baserat på BKR, Boverkets konstruktionsregler. ISverige dimensioneras dammar för att motstå de två globala brottmoderna glidningoch stjälpning. Frågan som behöver besvaras är om det finns eller kan finnas någrakombinationer av brottmoder som borde beaktas vid dimensionering av dammar.2009 antogs Eurokoderna och trädde i kraft 2011. Eurokoderna har ersatt BKR viddimensionering av de flesta konstruktioner i Sverige. I Eurokod används partialko-efficienter för att garantera säkra konstruktioner.

Syftet med denna rapport var att undersöka om dagens brottkriterium är tillräck-liga för att beskriva hur dammar går till brott. Rapporten behandlar även möj-ligheten att övergå från att dimensionera dammar enligt RIDAS till att dimen-sionera enligt Eurokod. För att besvara detta genomfördes en litteraturstudie avrapporterade dammbrott. Dessutom genomfördes analytiska beräkningar för fleraolika typer av dammar med varierande geometri och lastfall. Programmen CADAMoch BRIGADE användes som ytterligare verktyg för att analysera brottmoderna.

Enligt resultat från de analytiska beräkningarna tillsammans med FE-beräkningarnaansågs limit turning inträffa och genererade i högre grad en lägre säkerhetsfaktoreri jämförelse med stjälpning. Limit turning kan förklars som delvis stjälpande ochinkluderar brott av bergmassan. Brottmodet är dock beroende av kvalitéten hosberget och det krävs undersökningar av grunden för att kunna göra en god uppskat-tning av dammens verkliga beteende.

Sammanställningen av tidigare brott visade att nu gällande brottkriterier är lämpligaoch troligtvis tillräckliga. Utmaningen är istället att säkerställa att konstruktion-erna är korrekt utförda och därmed uppfyller dessa brottkriterier. En övergångtill Eurokod tycks vara möjlig för de globala brottmoderna, dock är det väsentligtatt partialkoefficienterna justeras för att uppnå resultat som överensstämmer medRIDAS, särskilt för stjälpning.

Keywords: Gravitationsdammar, betong, brottkriterium, RIDAS, Eurokod, limitturning

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Preface

This thesis was carried out from January to June 2015 at SWECO Energuide ABin collaboration with the Division of Concrete Structures, Department of civil andArchitectural Engineering at the Royal Institute of Technology (KTH). The projectwas initiated by Dr. Richard Malm, who also supervised the project, together withPh.D. candidate Daniel Eriksson and Adjunct Prof. Erik Nordström.

We would especially like to thank Richard Malm for the continuous support whichhas been a great encouragement. We would also like to thank Daniel Eriksson foralways finding time to help and guide us throughout this project. We also wish tothank Erik Nordstöm for his guidance and advice.

We would like to thank the division at SWECO Eneriguide AB for their warmwelcome and for an inspiring job environment. We would like to give special thanksto Johan Nilsson for the support and help during our work at SWECO.

Stockholm, June 2015

Lisa Broberg & Malin Thorwid

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Contents

Abstract iii

Sammanfattning v

Preface vii

1 Introduction 11.1 Background . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 11.2 Aim of report . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 21.3 Limitations . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 21.4 Structure of report . . . . . . . . . . . . . . . . . . . . . . . . . . . . 3

2 Concrete gravity dams 52.1 Gravity dams . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6

2.1.1 Massive dams . . . . . . . . . . . . . . . . . . . . . . . . . . . 62.1.2 Buttress dams . . . . . . . . . . . . . . . . . . . . . . . . . . . 72.1.3 Gate section . . . . . . . . . . . . . . . . . . . . . . . . . . . . 82.1.4 Support methods . . . . . . . . . . . . . . . . . . . . . . . . . 10

2.2 Stability analyses . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 102.2.1 Design loads . . . . . . . . . . . . . . . . . . . . . . . . . . . . 102.2.2 Failure modes . . . . . . . . . . . . . . . . . . . . . . . . . . . 11

3 Methods for stability analyses 153.1 RIDAS . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 15

3.1.1 Design loads . . . . . . . . . . . . . . . . . . . . . . . . . . . . 163.1.2 Failure modes . . . . . . . . . . . . . . . . . . . . . . . . . . . 20

3.2 Eurocode . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 213.2.1 Design loads . . . . . . . . . . . . . . . . . . . . . . . . . . . . 223.2.2 Failure modes . . . . . . . . . . . . . . . . . . . . . . . . . . . 27

3.3 Limit turning . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 283.3.1 Crushing resistance . . . . . . . . . . . . . . . . . . . . . . . . 283.3.2 Failure criteria . . . . . . . . . . . . . . . . . . . . . . . . . . 30

4 Failure modes of concrete dams 334.1 Documentation of failures . . . . . . . . . . . . . . . . . . . . . . . . 334.2 Compiled failures . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 34

4.2.1 Comparison of properties . . . . . . . . . . . . . . . . . . . . . 354.2.2 Failure type . . . . . . . . . . . . . . . . . . . . . . . . . . . . 36

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4.2.3 Failure mode . . . . . . . . . . . . . . . . . . . . . . . . . . . 374.3 Description of failures . . . . . . . . . . . . . . . . . . . . . . . . . . 38

4.3.1 Documentation regarding failures . . . . . . . . . . . . . . . . 394.4 Results of the compiled failures . . . . . . . . . . . . . . . . . . . . . 49

5 Stability analyses 535.1 Studied dams . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 53

5.1.1 Input data . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 545.1.2 Previously studied dams . . . . . . . . . . . . . . . . . . . . . 58

5.2 Stability Calculations . . . . . . . . . . . . . . . . . . . . . . . . . . . 595.2.1 Design approaches . . . . . . . . . . . . . . . . . . . . . . . . 595.2.2 Parametric study . . . . . . . . . . . . . . . . . . . . . . . . . 62

5.3 CADAM . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 625.3.1 Stability calculations . . . . . . . . . . . . . . . . . . . . . . . 635.3.2 Modelling . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 64

5.4 FE-analysis . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 655.4.1 Studied dams . . . . . . . . . . . . . . . . . . . . . . . . . . . 665.4.2 Model definition . . . . . . . . . . . . . . . . . . . . . . . . . . 67

6 Results and discussion 716.1 Analytical analyses . . . . . . . . . . . . . . . . . . . . . . . . . . . . 71

6.1.1 Design approaches . . . . . . . . . . . . . . . . . . . . . . . . 716.1.2 Parametric study . . . . . . . . . . . . . . . . . . . . . . . . . 796.1.3 Previously studied monoliths . . . . . . . . . . . . . . . . . . . 84

6.2 Analyses of limit turning . . . . . . . . . . . . . . . . . . . . . . . . . 846.2.1 Analytical analysis . . . . . . . . . . . . . . . . . . . . . . . . 846.2.2 FE-analysis . . . . . . . . . . . . . . . . . . . . . . . . . . . . 87

7 Conclusions 917.1 Failure modes of concrete dams . . . . . . . . . . . . . . . . . . . . . 917.2 Analytical calculations . . . . . . . . . . . . . . . . . . . . . . . . . . 927.3 Design guidelines . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 937.4 Future studies . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 94

Bibliography 95

A Compiled failures 101

B Results analytical analyses 103

C Output values for dams 105

Chapter 1

Introduction

1.1 Background

Most dams in Sweden were built during 1950 to 1960 on solid good quality rock.The dams were built under different conditions and safety regulations compared tothe demands stated today (Andersson, 2012). The knowledge of rock mechanics andmaterial properties of concrete along with the building techniques have improved.Today the construction of new dams in Sweden is limited by regulations concerningthe preservation of the environment. Therefore, the design of dams mostly involvesmaintenance and reparation of existing dams.

Knowledge of why and how dam failure occurs, may help prevent or minimise thedamage. The indication of how a dam behaves prior to failure is therefore of greatimportance in order to prevent failure. It is also important since it provides guidanceon how to monitor and measure dams, what types of sensors and where these sensorsshould be placed to get early indications of possible failures. Risk and safety areessential in dam design due to the radical consequences a failure would cause to thesurroundings, as seen in Figure 1.1.

Figure 1.1: Baldwin Hills Reservoir after the disaster 1963 (Wilson, 1963).

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CHAPTER 1. INTRODUCTION

The consequences of failure could in the worst case scenario lead to lives lost andeconomic damage. Therefore, the government through the public utility SvenskaKraftnät, stated new requirements concerning higher safety demands for the existingdams (SFS 2014:114). The new requirements also concern the classification of thedams in Sweden; all dams must be classified, if a failure could result in severeconsequences.

In Sweden, the dam owner is responsible in the event of a failure or an accident.The Swedish power companies have issued the Swedish power companies’ guidelinesfor dam safety, RIDAS, based on the construction rules BKR (2010), the NationalBoard of Housing, Building and Planning. Since 2011, the Eurocodes (the Europeanconstruction standards) together with EKS 9 (2013), have replaced BKR in Sweden.However, Eurocode does not account for the design of dams (Andersson, 2014).

Today Swedish dams are designed to withstand two global failure modes; slidingand overturning of the entire monolith. There are questions regarding if there areor can be any combinations of the failure modes, that should be considered in thedesign of concrete dams.

1.2 Aim of report

The main focus of this report is to analyse the different types of failures that haveoccurred and can occur in concrete gravity dams, by examining the influence ofdifferent factors. The aim of this study is to answer the following questions:

• Is analytical calculations based on the global failure modes: sliding and over-turning enough to describe the failure of the dam? Are there other potentialfailure modes not covered by these analytical calculations?• Is a transition to Eurocode possible for dam design? Are the design guidelines

according to Eurocode comparable to RIDAS?

The stated questions will be answered by performing a literature study of reporteddam failures and analytical calculations for several concrete gravity dams. In addi-tion a FE-analysis will be performed for comparison.

1.3 Limitations

This report only include concrete gravity dams and further limitations for the litera-ture study are stated in Chapter 4. The analytical calculations only concern dams onrock foundation. The analyses are limited to Swedish conditions regarding materialproperties and design parameters. The influence from seismic loads is not includeddue to that it is not considered for design in Sweden, while it internationally maybe of great importance for the design.

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1.4. STRUCTURE OF REPORT

1.4 Structure of report

Chapter 2 includes a presentation of theory behind the key concepts of concretegravity dams. The main features for stability analyses of concrete gravity damsare presented. The different design loads for the stability analysis are stated andillustrated. A brief presentation of the causes of concrete gravity dam failure byexplanation of the failure modes is presented.

Stability analyses according to the guidelines RIDAS and Eurocode are presentedin Chapter 3. How the guidelines account for the design loads and explanations ofthe analytical calculation methods for the failure modes described in Chapter 2 aregiven.

A compilation of reported failures including causes and failure modes is presentedin Chapter 4. Already known facts about the presented failures of concrete damsare compiled and the sources of failures that has occurred are detected.

The analytical analyses in Chapter 5 describe the stability calculations for severaldifferent dams with varying geometry and loading conditions. A parameter study isused to determine the most influential parameter and to adapt the design of damsaccording to Eurocode with the stability calculations in agreement with RIDAS.The program CADAM and the software BRIGADE, are used as tools to enablea comparison to the analytical calculations. The key concepts on how to performstability analyses with these tools are presented in this chapter. The results fromthe analyses described in Chapter 5, are presented in Chapter 6.

In Chapter 7, the conclusions of the different concrete dam failure analyses arepresented.

Appendix A includes the full compilation about the failures of concrete dams pre-sented in Chapter 4. Appendix B includes the results from the analytical calculationsdescribed in Chapter 5. The loads and level arms from the analytical calculationsare presented in Appendix C.

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Chapter 2

Concrete gravity dams

There are different types of concrete dams, which are distinguished by how thewater pressure of the reservoir is transferred down to the ground. Descriptions ofthe different types of concrete gravity dams are included in this chapter. The twomost common types are massive and buttress dams, presented in Section 2.1.1 andSection 2.1.2 respectively. Gate sections mainly consist of pillars and spillways, bothare of gravity dam type and are described in Section 2.1.3. Figure 2.1, shows thedifferent dam types included in this report.

Figure 2.1: The different types of dams included in this report, a) spillway and pillar,b) massive and c) buttress.

For other types of dams, not included in this report, concrete can also be used in em-bankment dams, as a central or upstream membrane, as retaining walls for spillwaysor used for many secondary functions. Embankment dams are usually associatedwith at least one concrete dam part, either intake and/or discharge facilities. Con-crete arch dams were introduced relatively late and therefore have a uniform, andsomewhat higher quality. Arch dams are founded on rock and are of slender typewith concaved arches (Kleivan et al., 1994).

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CHAPTER 2. CONCRETE GRAVITY DAMS

2.1 Gravity dams

2.1.1 Massive dams

Massive or gravity dams are solid structures, designed to resist the external forcesby its dead weight. Today, gravity dams are mainly constructed with concrete,compared to the previously used method of stone masonry. The development of newconcrete gravity dams is ongoing and the Roller-Compacted Concrete dam (RCC)is an example of that. The RCC dam has a limited use of formwork, consists of adrier mix and is placed in a manner similar to paving, i.e. compacted with vibratingrollers. The benefits are cost beneficial with simple faster construction. The damsare built with no joints or reinforcement, with low cement content and the use of flyash that enable less heat generation while curing (Kleivan et al., 1994).

Solid concrete structures maintain stability against loads due to the geometric shape,mass and strength of the concrete (Ali, 2012). A gravity dam consists of either acontinuous or a series of concrete monoliths separated by expansion joints (RIDAS,2011). The monolith cross section is, in principle, triangular with a dam head, aninclination of the downstream face and a vertical upstream face, which also can havea small inclination, see Figure 2.2. The benefits of concrete gravity dams are thatthey are easily constructed on sites with a foundation of sufficient strength to carrythe weight of the dam (Ali, 2012).

Figure 2.2: Typical cross-section of a massive dam, reproduced from Bergh (2014).

The monoliths are mainly placed in a straight line or sometimes slightly curved, seeFigure 2.3 and are usually of a width between 5 m to 10 m (Bergh, 2014).

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2.1. GRAVITY DAMS

Figure 2.3: Stadsforsen, massive dam in Sweden (Malm, 2015).

2.1.2 Buttress dams

Over time, there has been a strong effort towards improving concrete quality. Therehas been a shift from the previously dominant gravity type dam such as massivedams, towards a more slender type of dam with reinforcement, known as a buttressdam (Kleivan et al., 1994).

Buttress dams consist of two rigidly connected elements, the upstream water barrier(frontplate) and the supporting buttress on the downstream side, which togetherform a monolith, see Figure 2.4. The upstream water barrier transfers the hydro-static pressure over to the buttress, which in turn transfers it down to the foundation.The water barrier is inclined so the vertical water loads, together with the weight ofthe concrete, act in favour for the stability of the monolith (DOI, 2009).

Figure 2.4: Cross-section of a buttress dam (Left), Section of a buttress dam (Right),reproduced from RIDAS (2011).

A buttress dam consists of a series of monoliths, connected by horizontal strutsacting as contraction joints, connecting the adjacent monoliths, see Figure 2.5. Thecasting arrangements and the construction are somewhat more demanding com-pared to most other types of dams. However, buttress dams are more suitable on

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weak foundations compared to gravity dams, due to the reduced volume of concrete(Bergh, 2014), while the contact pressure between the buttress and the foundationis considerably higher.

Figure 2.5: Rätan, buttress dam in Sweden (Vattenkraft.info, 2009).

2.1.3 Gate section

Concrete gravity dams also consist of gate sections to transport water in specificdirections and release water pressure on the dam structure. Concrete functions asa fastener for many different types of gate installations, with variable functions.The gate type could be sliding, roller or radial gates, with the function of either anoutlet gate, intake or daft tube (Kleivan et al., 1994). An example of a gate sectionin Sweden is shown in Figure 2.6

Figure 2.6: Lima, Spillway dam with two river type gates (Norconsult, 2012).

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2.1. GRAVITY DAMS

For a spillway section see Figure 2.7, where the overflow is designed with a verticalupstream face. The water is able to flow over the crest along the inclined down-stream side with training walls, keeping the water in place and finally the waterreaches the energy dissipating structure, forming a hydraulic jump to avoid erosionof the riverbed. Pillars, non-overflowing blocks function as enclosures of a number ofoverflow sections, see Figure 2.7. Usually spillway sections have gates and typically,radial gates see Figure 2.7.

Figure 2.7: Spillway section (Left), plane view of spillway (Right), reproduced fromRIDAS (2011).

For hydropower structures, the intake is the connection between the reservoir andthe waterway connected to the turbine of the hydroelectric power plant. Intakegates are normally designed with a trash rack preventing debris, ice, fish, etc. fromentering the intake. In addition, it also consists of a gate of river or tunnel type, toclose of the conduit, see Figure 2.8.

Figure 2.8: Section of a typical inlet and power station, reproduced from RIDAS(2011).

The behaviour of a spillway, discharge part, and intake part of the dam is similarto concrete gravity dams, thus the geometry and loading conditions may be morecomplicated (Westberg and Hassanzadeh, 2007). The surface of the concrete issubjected to very high water velocities as well as abrasion, and therefore must besteel plated.

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2.1.4 Support methods

There are different support methods for concrete gravity dams. Common methodsare to secure the dam body to the foundation by rock bolts or tendons. Rockbolts are non-pre-stressed reinforcement bars installed in the interface between thefoundation and the dam body. There are different types of fastners for the rockbolts and they can be secured to the foundation through anchors, cables, dowles orby grouting. The bolts are anchored in the dam body by adhesion. Rock tendonsconsist of pre-stressed cables or rods, anchorage and corrosion-inhibiting coating.The tendons can be unbonded or bonded to the surrounding concrete. The tendonis inserted into a casing and grouted after the tendon is stressed (PTI, 2000).

Another common method used to support concrete dams is earth support fill on thedownstream side of the dam, giving rise to stabilising forces. A grout curtain helpsto stabilise the dam by decreasing the uplift pressure. The grout curtain is achievedby inserting cement into pores and cavities in the ground. The grout curtain mightdeteriorate over time and therefore the decreased uplift is usually not accounted forin the design of dams (Ferc Engineering Guidelines, 2002). The uplift pressure couldalso be decreased by inserting drainage pipes.

2.2 Stability analyses

Concrete dams are massive large structures since they are designed to fulfil therequirements for stability. The design should also fulfil requirements for long lifespans and water tightness to withstand the permanent water pressure (Bond, 2014).The safety of the concrete dam is assured by designing according to failure criteria,for all combinations of loads using safety factors. The safety criterion is the definitionof the stress level when failure occurs. The safety factors are chosen to provide forall underlying uncertainties. Their magnitude should reflect the probability of theoccurrence for the particular load, the accuracy of conditions and the method ofanalysis. The factor of safety is thereby higher for foundation studies, because ofthe greater amount of uncertainty in assessing the load-resistance capacity of thefoundation.

2.2.1 Design loads

The loads included in the stability analysis should represent the actual loads actingon the concrete dam during operation. Many of the loads are unable to be exactlydetermined, the engineer is then responsible for estimating these loads based onavailable measurements, judgement and experience (Ferc Engineering Guidelines,2002).

The required loads acting on a dam for a stability analysis are shown in Figure 2.9.

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2.2. STABILITY ANALYSES

Figure 2.9: Loads acting on a dam, reproduced from (Bergh, 2014).

1. Hydrostatic pressure (P1-P2) – depends on the water level in the dam

2. Tailwater pressure (P3-P4) – depends on the tailwater level

3. Uplift pressure (P5) – hydrostatic pressure acting vertically, assumed to varylinearly from hydrostatic pressure at the heel to the tailwater pressure at thetoe

4. Dead weight (P6) – the weight of the concrete

5. Ice pressure (P7) – load acting on the face of the dam due to an ice cover

6. Silt pressure (P8) – settled sediments exerting active pressure towards the dam

7. Seismic loads (P9-P11) – horizontal and vertical accelerations caused by earthquakes

2.2.2 Failure modes

The definition of dam failure can differ between individuals, the general definitioncould be expressed as: "Collapse or movement of part of a dam or its foundation,so that the dam cannot retain water” (ICOLD, 1995).

Concrete dams have various failure behaviour. Sliding or shear failure is the mostcommon failure for dams constructed on rock. The dam may fail due to crushing,i.e. the failure of its materials when the compressive stresses exceed the acceptablestresses. Concrete cannot withstand sustained tensile stress and if the tension thatdevelops in the concrete exceed its tensile strength, it could lead to ultimate failure.The dam may also fail due to overturning where it rotates about the toe (Ali, 2012).

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Overturning

Overturning occurs when the forces acting on the dam causes rotation of the dam,see Figure 2.10. Overturning is analysed by calculating a factor of safety, whichis defined as the ratio of stabilising and overturning moment. These moments arecalculated around its toe or another weak point in the structure or foundation. Forthe overturning failure, it is also important that the resultant is located in the midthird of the base area since this will assure that the whole base of the dam is undercompression. If tensile stress can be avoided it will reduce crack propagation inthe concrete. The criterion is verified by application of the Navier equation for acantilever action under combined axial and bending load (Bergh, 2014).

Figure 2.10: Overturning failure around the dam toe.

Sliding

Sliding occurs when the horizontal forces exceed the frictional resistance. Slidingcan be divided in to three different kinds of failures (Gustafsson et al., 2008):

1. Failure in the interface between the concrete and the foundation(Figure 2.11 a).

2. Failure in weak planes of the foundation, such as cracks (Figure 2.11 b).

3. Failure in the solid foundation (Figure 2.11 c).

Figure 2.11: Different types of sliding failures, reproduced from Gustafsson et al.(2008).

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2.2. STABILITY ANALYSES

There are different views on whether cohesion between the concrete and the foun-dation can be accounted for in the sliding stability. The reason is that it is difficultto quantify through borings’ and testing. Higher allowable safety factors may beapplied, if cohesion is included in the calculations for sliding stability (Ferc Engineer-ing Guidelines, 2002). If cohesion is not accounted for, the concrete-rock interfaceis treated as unbonded giving a conservative method that may result in expensiveand over-strong structures (Krounis, 2013).

Failure in the interface between the foundation and the dam body is normally ac-counted for in the design of dams. Though failure more often occurs in weak planesof the foundation, see Figure 2.11 (DOI, 2012). Sliding can also occur in the dambody at weak planes such as lift joints or along cracks, this failure is seldom analysedexcept for high dams (Ali, 2012).

Limit turning

According to Fishman (2007), the classical failure modes sliding and overturning,do not account for material failure. Classical overturning failure is unrealistic asit requires infinitely strong rock and concrete. Fishman infers that the one failuremode, either limit turning or sliding, giving the lowest stability factor should beused for the design of the structure and decisions regarding interface preparation(Fishman, 2009).

Fishman states that the stresses developing below the upstream side will result ina tensile crack along the rock, see Figure 2.12. A compressive zone will be formedin the rock, underneath the toe, due to the applied forces on the dam. When thestresses exceed the crushing resistance of the rock, a crushing zone is formed. Thesize of the crushing zone depends on the strength of the rock, for a weaker rock thecrushing zone will extend further to the upstream side. The turn axis appears wherethe tensile crack and the compressive crack meet. The concrete and rock will act asa single body and failure will occur when they rotate about the new rotation point.This failure mechanism is called limit turning, which is similar to the overturningmethod although it also accounts for the strength of the rock. The result gives amore reliable and conservative safety factor (Fishman, 2009).

Figure 2.12: Limit turning failure.

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Chapter 3

Methods for stability analyses

In this chapter, the design methods for stability analyses are presented. Thereare different methods for analysing failure of dams, varying between countries ormethods. In Sweden, RIDAS (2011), is used for the design of dams and is basedon BKR. 2011 BKR was replaced with EKS 9 (2013). Today, Eurocode is used andfunction as guidelines providing a common structural design in European countries.

A revision of RIDAS, to incorporate Eurocode instead of BKR has begun. Thetransition work is led by Svensk Energi, responsible for RIDAS. The research com-pany, Energiforsk, previously known as Elforsk until January 2015, has also startedto work with the transition from BKR to Eurocode. The investigations has so farmostly been focused on cross-section design (Andersson, 2014).

For the stability analyses, there is an ongoing project financed by Energiforsk, wherea structural reliability based method is under development. In this report, thefocus will be on RIDAS, using the deterministic method, compared to Eurocodessemi-probabilistic method, resulting in the use of partial factors, for calculations ofstability (Westberg, 2014).

The denotations from RIDAS and Eurocode were used in Section 3.1 and Section3.2, respectively.

3.1 RIDAS

RIDAS (2011), is based on BKR, with adjustments for specific requirements for con-crete dams. BKR, as mentioned before, is not valid today however it may still beused if the contents do not conflict with the Eurocodes (Andersson, 2014). Guide-lines are given for the design of dams together with control and reconstruction ofexisting dams. RIDAS states requirements for stability, strength and durability ofthe dam and what criteria to fulfil. The requirements and criteria concern gravitydams, where RIDAS have included the most common types; massive and buttressdams, including spillway, inlet dams and pillars.

15

CHAPTER 3. METHODS FOR STABILITY ANALYSES

3.1.1 Design loads

RIDAS (2011) includes guidelines to determine design loads acting on concrete dams.It states how to account for the loads presented below. RIDAS also gives guidancehow to account for rock anchors (described below), temperature effects, creep andshrinkage, and traffic loads (included if unfavourable), which are not described inthis report.

Dead weight

For design of new concrete dams, the dead weight for reinforced concrete is assumedto be 23 kN/m3 if no material tests are available. For existing dams, the dead weightshould be determined from material tests or from information about the design.

Hydrostatic pressure

Both the water pressure on the up- and downstream side should be accounted for.The most unfavourable combinations of up- and downstream water levels applied tothe dam determine the water pressure to be used in the calculations.

Uplift pressure

The uplift pressure distribution varies for different dam types and designs with orwithout drain pipes. For massive dam structures without drainage where the wholefoundation area is under compression, the uplift pressure distribution varies linearlyfrom the upstream to the downstream side.

For massive dams the uplift pressure can be reduced by the use of drain pipes asshown in Figure 3.1 and Figure 3.2.

16

3.1. RIDAS

Figure 3.1: Uplift distribution for a dam with a drainage pipe in the rock and adrainage tunnel by the rock surface, reproduced from RIDAS (2011).

The uplift distribution in Figure 3.1 will vary from H to 0.3 · (H − h) + h closestto the drainage tunnel and varies linearly to h at the toe of the monolith, with nouplift beneath the drainage tunnel. H is the headwater level and h is the tailwaterlevel.

Figure 3.2: Uplift distribution for a dam with a drainage pipe in the rock anddrainage tunnel in the concrete, reproduced from RIDAS (2011).

The uplift distribution in Figure 3.2 will vary from H to 0.5 · (H − h) + h closestto the drainage tunnel and varies linearly to h at the toe of the monolith. H is theheadwater level and h is the tailwater level.

17


For buttress dams, the uplift distribution is assumed to vary linearly over thethickness of the frontplate. If the buttress is thicker than 2 m, the uplift pressureunderneath the buttress should be included as shown in Figure 3.3.

Figure 3.3: Distribution of uplift pressure for a buttress dam with a buttress thickerthan 2m, reproduced from RIDAS (2011).

For spillways, the uplift distribution is assumed similar to massive dams. The upliftdistribution in Figure 3.4 is assumed for pillars, where w is the width of the pillar.The uplift varies from full uplift pressure to zero at the distance w from the spillway.

Figure 3.4: Uplift distribution for pillars, reproduced from RIDAS (2011).

The effect from cement grouting is not considered in the uplift pressure distribution,due to the strength of the cement decrease with time due to deterioration. The groutcurtain is only considered as extra safety and should not be accounted for unlessre-grouting is possible. This is seldom the case due to difficulties incorporatingre-grouting tunnels in the dam body design.

18

3.1. RIDAS

Ice load

The intensity of the horizontal ice pressure depends on the geographic location,altitude and local conditions for the dam. RIDAS suggest the horizontal ice pressure50 kN/m with an ice thickness of 0.6 m, for dams located in the southern part ofSweden. Dams located in the middle part, should be designed for an ice load of100 kN/m with an ice thickness of 0.6 m. An ice load of 200 kN/m with an icethickness of 1 m, is suggested for the rest of Sweden. The resultant of the icepressure is assumed to be located at one third of the ice thickness, calculated fromthe top surface of the ice.

Rock anchors

For lower dams, it can be hard to achieve stability, and according to RIDAS in thesecases it is allowed to assign a load capacity of 140 MPa to the rock bolts. This isapplied for dams that have a headwater level less than 5 m and do not belong toany of the two highest safety classes.

For all other dams, rock anchors should not be considered in the stability calcu-lations, due to the complications of verifying their strength. However, it is statedthat the installation of rock anchors of the dimension φ25− 32 is a good preventivemeasure.

Earth pressure

Soil may be added as downstream support fill to increase stability. The earth pres-sure should be determined as a at-rest pressure and it should be calculated as thelowest theoretical pressure that may occur. The soil density and earth pressure co-efficient should be obtained from in-situ tests. If testing is not possible, the valuesin Table 3.1 may be used.

Table 3.1: Example values for unit weight and coefficients for earth pressure RIDAS(2011).

MaterialUnit weight density [ kN/m3] Friction angle [ ◦] Coefficient for earth pressure [-]Un-saturated Saturated φ At-rest K0 Active Ka

Rockfill 17.5 11 42 0.33 0.20Gravel 18 11 35 0.43 0.27Sand 18 11 32 0.47 0.31Moraine 21 13 34 0.45 0.29

19


3.1.2 Failure modes

According to RIDAS (2011) there are three failure modes that need to be analysedfor stability; sliding, overturning and the bearing capacity of the concrete and thefoundation.

Dam stability analyses are performed using safety factors for overturning and allow-able friction coefficients for sliding, to achieve a safe design.

RIDAS has listed different load combinations to be analysed, these are divided into; normal load combinations, exceptional load combinations and accidental loadcombinations. The loads are calculated without partial factors and are analysed forindividual monoliths.

Overturning

For overturning the requirement of the safety factor s is defined according toTable 3.2.

Table 3.2: Saftey factor for overturning.

Load case Safety factor (s)

Normal 1.50Exceptional 1.35Accidental 1.10

The safety factor defines the relation between stabilising and overturning moment,see Equation (3.1), and should not be lower than the safety factor in Table 3.2.

s =Mstab

Mover

(3.1)

Sliding

RIDAS states that sliding should be analysed between the interface of the concreteand in the foundation, along potential weak planes and in weak points in the dambody. Stability against sliding is achieved if the sum of the horizontal forces dividedby the vertical forces, see Equation (3.2), does not exceed the maximum allowedfriction coefficient, see Table 3.3.

µ =RH

RV

≤ µmax =tan δgsg

(3.2)

20

3.2. EUROCODE

where

RH is the resultant of the horizontal forces.Rv is the resultant of the vertical forces.tan δg is the friction angle.sg is the safety factor.

Table 3.3: Maximum friction coefficient, µmax

Foundation Normal loadcase

Exceptionalload case

Accidentalload case

Rock 0.75 0.90 0.95Gravel, sand and moraine 0.5 0.55 0.60Silt 0.40 0.45 0.50

The maximum friction coefficient can be calculated according to Equation (3.2)where the values for the safety factor sg are presented in Table 3.4.

Table 3.4: Saftey factors sg for calculations of µmax.

Foundation Normal loadcase

Exceptionalload case

Accidentalload case

Rock 1.35 1.10 1.05Gravel, sand and moraine 1.50 1.35 1.25Silt 1.50 1.35 1.25

Cohesion between the dam and the foundation should not be considered in thecalculations for the resistance against sliding according to RIDAS.

When calculating stability against sliding, the value for the maximum allowed fric-tion coefficient (µmax) is calculated according to Equation (3.2), with values of thesafety factor from Table 3.4 and the friction angle from geotechnical investigations.The values in Table 3.3 can be used for dams constructed on a foundation of goodquality, when calculating stability against sliding.

3.2 Eurocode

Eurocode is the European standard for technical rules in construction work, pro-viding a common structural design tool in European countries. Eurocode clearlystates that their guidelines do not cover the design of dams. This is due to the highsafety required for dams and that other aspects than for usual design need to beconsidered. This section is therefore solely based on the authors assumptions onhow to apply Eurocode to dam design. Therefore, in this report a compilation ofinformation from the listed Eurocodes below was performed in order to obtain amethod applicable for stability analyses of dams.

21


• EN 1990 “Basis for structural design”.• EN 1991 “Actions on structures”.• EN 1992 “Design of concrete structures”.• EN 1997 “Geotechnical design”.

The same failure modes as described in Section 2.2.2 were analysed. The values andequations in the following sections were based on the assumption that dams can beconsidered as comparable with retaining wall structures.

3.2.1 Design loads

The safety is applied on the loads by partial factors. The load acting on the structureis multiplied with the partial factor γ to define the design load. The loads areclassified according to variation in time: permanent or variable loads and whetherthe load is favourable or unfavourable, which would result in different partial factors.

When designing geotechnical structures, different approaches are used. In accor-dance with retaining wall structures, the concrete dams were assigned the designapproach 3 (DA 3). The different approaches give different values for the partialfactors; for load and load effects, soil parameters and the strength (EC 7, 2011).

Design approach 3 states that different partial factors should be used for geotech-nical actions and structural actions. Geotechnical actions are defined as actionstransmitted to the structure by the ground, fill, standing water or groundwater.For structural actions the strength of the material is significant. For the structuralactions Equation (3.3) and (3.4) are used. For geotechnical actions, Equation (3.5)is used (EC 0, 2002). The partial factors for the equations are listed below in Table3.5.

Ed = Σγd · γG ·Gk + Σγd · γQ · ψ0,i ·Qk,i (3.3)

Ed = Σγd · ξ · γG ·Gk + γd · γQ,1 ·Qk,1 + Σγd · γQ,i · ψ0,i ·Qk,i (3.4)

Ed = Σγd · γG ·Gk + γd · γQ,1 ·Qk,1 + Σγd · γQ,i · ψ0,i ·Qk,i (3.5)

where

γG is the partial factor for permanent actions.γd is the partial factor depending on safety class.Gkg is the characteristic value of a permanent action.γQ,i is the partial factor for variable action.ψ0,i is the factor for combination value of a variable action.Qk,i is the characteristic value of a single variable action.ξ is the reduction factor.

22

3.2. EUROCODE

Table 3.5: Partial factors according to Eurocode (EC 7, 2011).

Load combination equation 3.3/3.4 3.5

For an unfavourable permanent load γG = 1.35 γG = 1.1For a favourable permanent load γG = 1.0 γG = 1.0For an unfavourable variable load γQ = 1.5 γQ = 1.4For a favourable variable load γQ = 0 γQ = 0Reduction factor ξ = 0.89/- -

Structures are classified into different safety classes depending on the harm a failurewould cause, the definitions are stated in Table 3.6. For calculations of stability,according to the partial factor method in EC 0 (2002), the partial factor γd is appliedand this value depends on the safety class of the structure. The partial factor forthe different safety classes is shown in Table 3.7.

Table 3.6: Consequence classes (EC 0, 2002).

Consequences class Description

CC3 High consequence for loss of human life, oreconomic, social or environmentalconsequences very great.

CC2 Medium consequence for loss of human life,economic, social or environmentalconsequences considerable.

CC1 Low consequence for loss of human life, andeconomic, social or environmentalconsequences small or negligible.

Table 3.7: Partial factors γd according to safety class (EC 0, 2002).

Safety class Partial factor, γd1 0.832 0.913 1.0

Dead weight

The dead weight stabilises the dam and therefore acts as a favourable, permanentload and is a structural action. For reinforced concrete with normal weight, thedensity γc= 24 kN/m3 should be used for calculations of the dead weight (EC 1,2013).

23


Hydrostatic pressure

The horizontal water load (HW) acting on the upstream face, shown in Figure 3.5,is a permanent and unfavourable load while horizontal tailwater load (TW) on thedownstream side and the vertical load (VW) are permanent and favourable loads.The water loads are categorised as geotechnical actions. For analyses where thehydrostatic pressure is increased above the headwater level, the hydrostatic pressurecan be classified as a variable load.

Figure 3.5: Hydrostatic pressure.

According to EC 1 (2013) the density for fresh water is set to γw = 10 kN/m3 andthe loads caused by water should be determined with respect to the water level.

No combination factor is given for the water load; therefore, in this report, a valuein the interval between the value for the highest snow load and the value for imposedloads on buildings was assumed. The combination factor ψ0,w= 0.75 can thereforebe used for variable water loads.

Uplift pressure

The vertical uplift pressure is considered as an unfavourable permanent load andgeotechnical action. The density of water and the partial factor can be set accordingto the hydrostatic pressure. An additional horizontal uplift pressure will be presentfor monoliths without horizontal bottom surfaces. The horizontal uplift pressure canact as both an favourable and an unfavourable load, depending which direction themonolith is inclined. The horizontal uplift pressure resultant can also, depending onthe location, vary between unfavourable and favourable for sliding and overturning.

Ice load

The ice load is an unfavourable variable load and is a structural action. In thisreport, the combination factor ψ0,ice = 0.8 was chosen for the load combinations,based upon the highest value for snow loads.

24

3.2. EUROCODE

Rock anchors

Rock anchors are considered as permanent favourable loads and geotechnical actions.The design strength of the reinforcing rock anchors may be calculated according toEquation (3.6).

fyd =fykγs

(3.6)

where

fyk is the characteristic strength of the reinforcement.γs is the partial factor for the reinforcement.

The partial factor applied to untensioned and tensioned reinforcement bars is in thisreport assumed to be γs = 1.15 based on EC 2 (2011).

Earth pressure

Earth pressure is both a favourable and an unfavourable permanent load and is inthis report considered as a geotechnical action. Eurocode states that the soil prop-erties should be chosen from investigations or by theoretical or empirical correlationor from other relevant documentation. If standard values from tables are used, thecharacteristic values should be chosen with great care. According to the design ofretaining wall structures, the determination of the earth pressure should be takenas at-rest pressure, if no movement of the wall relative the ground takes place. Thelateral earth pressure coefficient, KO is calculated according to Equation (3.7) fora horizontal backfill and according to Equation (3.8) for an inclined backfill anddepend on the friction angle (EC 7, 2011).

Horizontal backfill:KO = (1− sinϕ′) ·

√OCR (3.7)

Inclined backfill:KO;β = KO · (1 + sin β) (3.8)

where

ϕ′ is the effective friction angle.OCR is the overconsolidation ratio.β is the slope of the soil.

Values for the unit weight density and friction angle in Table 3.8, was obtained fromTrafikverket (2011).

25


Table 3.8: Material properties for ground materials.

MaterialUnit weight density [ kN/m3] Friction angle [ ◦]Un-saturated Saturated ϕ′

Rockfill 18 11 45Gravel 19 12 37Sand 18 10 35Gravelly moraine 20 13 38Sandy moraine 20 12 35Silty moraine 20 11 33

Eurocode also states that the earth pressure should be calculated according to thechosen design approach, as shown in Equation (3.9). The design value of the earthpressure is:

Xd =Xk

γM(3.9)

where

Xk is the characteristic value of the material property.γM is the partial factor of the material property.

The partial factor for material properties was chosen in accordance with designapproach (DA3) to γM = 1.3. In most cases the earth pressure acts as a stabilisingforce, leading to a decreased force, resulting in a more conservative value for theearth pressure. If the earth pressure is active, the diagram in Figure 3.6 can be usedto obtain the active earth pressure coefficient, Ka.

26

3.2. EUROCODE

SS-EN 1997-1:2005 (Sv)

128

(3) Värdena på de effektiva jordtryckskoefficienterna kan hämtas från Figurerna C.1.1 till C.1.4 för Ka och C.2.1 till C.2.4 för Kp.

(4) Alternativt kan den analytiska metoden, som beskrivs i C.2, användas.

(5) I skiktade jordar bör koefficienterna K normalt endast bestämmas av skjuvhållfasthetsparametrarna på djupet z, oberoende av värdena på andra djup.

Figur C.1.1 – Aktiva, effektiva jordtryckskoefficienter Ka (den horisontella delen): stöttad horisontell markyta (� = 0)

SIS fleranvändarlicens: SWECO Sverige AB. 2013-02-22

Uppdatering enligt EKS9 har gjorts av Anette Sjölund och Elizaveta Pronina. Senaste revidering 2015-04-16.Tillägg och kommentarer i detta dokument har gjorts av Emma Persson. Teknikområde Grundläggning. Senaste revidering 2013-02-25.

Figure 3.6: Active earth pressure (EC 7, 2011).

3.2.2 Failure modes

EC 7 (2011) defines how to perform the design of geotechnical structures. Thecalculation model should describe the behaviour of the foundation and be reliableand give an error on the safe side.

It should be verified that ultimate limit state is not exceeded for:

• Internal failure or excessive deformation of the structure or structural elements,in which the strength of the structural material is significant in providingresistance (STR).• Failure or excessive deformation of the ground, in which the strength of soil

or rock is significant in providing resistance (GEO).

Overturning

The failure criterion presented in Equation (3.10) for ultimate limit state fromEC 0 (2002), defines the safety against overturning as:

Md,dst ≤Md,stb + Td (3.10)

27


where

Mdst is the overturning moments.Mstb is the stabilising moments.Td is the shearing resistance.

If the shearing resistance, Td, is included, it should not have a considerable effect onthe result.

Sliding

The failure criterion presented in Equation (3.11) for ultimate limit state accordingto EC 7 (2011), defines the safety against sliding as:

Hd ≤ Rd +Rp;d (3.11)

where

Hd is the design value of unfavourable horizontal forces.Rp;d is the design value of favourable horizontal forcesRd is the design shear resistance.

The design shear resistance is calculated by Equation (3.12).

Rd = V ′d ·tan δdγM

(3.12)

where

V ′d is the design value of the effective vertical load.tan δd is the design friction angle.

According to Eurocode, the friction angle tan δd, should be determined based ongeotechnical investigations.

3.3 Limit turning

3.3.1 Crushing resistance

An important parameter in the limit turning calculations is the crushing resistanceof the rock mass, Rcr. This is a better estimation of the resistance to shear loadingcompared to the shear strength parameters friction angle and cohesion. The crushingresistance of the rock mass should be obtained from geotechnical investigations anddepends on the peak shear and normal stresses acting on the rock. When thecrushing resistance is exceeded and the crushing zone is formed, a limit turning

28

3.3. LIMIT TURNING

failure will occur (Fishman, 2007). From experiments performed by Fishman therelationship between the crushing resistance and uniaxial stress was obtained, shownin Equation (3.13).

Rcr = 1.47 · σc (3.13)

If no investigations are available, values from Table 3.9 may be used in the calcula-tions (Fishman, 2009).

Table 3.9: Crushing resistance Rcr for different categories of rock mass (Fishman,2009).

Category of rock Type of foundation Parameter Rcr (MPa)

I Massive, large fragmental, laminated, platy, verylow and low jointed, unweathered rockcharacterised by uniaxial compression strength in asample σc > 50 [MPa]

20.0

II Medium jointed, inconsiderably weathered rockcharacterised by σc > 50MPa

10.0

III Intensively jointed rock with σc = 15− 50MPa andinconsiderably weathered and low jointed rock withσc = 5− 15Mpa

5.0

IV Semi-rock, platy, thin-platy, medium, high and veryhigh jointed with σc = 5MPa

2.5

29


3.3.2 Failure criteria

The stability factor is the ratio between the sum of resisting moments and the sumof turning moments. Including the moment from the force of the peak crushingresistance S. The moments are calculated relative to the turning axis O, as sown inFigure 3.7.

Figure 3.7: Principles of limit turning, reproduced from Fishman (2007).

The position of O axis is determined as follows:

O = (a, d) = (N

t ·Rcr

, [(h2 + 2 · a · e− a2)1/2 − h]) (3.14)

The force of peak crushing resistance is defined in Equation (3.15).

S = (a2 + d2)0.5 · t ·Rcr (3.15)

The moment of the peak crushing resistance will be calculated about the O axis andadded to the resisting moment:

Mp.c = S · bcr · 0.5 (3.16)

Limit turning stability factor, relative to the O axis:

Fs =ΣMr

ΣMt

(3.17)

30

3.3. LIMIT TURNING

where

a is the x-distance from the downstream toe B to the position of the turningaxis O.

d is the y-distance from the downstream toe B to the position of the turningaxis O.

N is the resultant of the vertical forces.T is the resultant of horizontal forces.t is the width of the structural section along a projected center-line or the

thickness of the buttress.Rcr is the crushing resistance of the rock.h is the lever arm of the horizontal forces T relative the downstream toe B.e is the lever arm of the vertical forces N relative the downstream toe B.bcr is the length of crushing plane OB.∑Mr is the sum of resisting moments.∑Mt is the sum of turning moments.

31

Chapter 4

Failure modes of concrete dams

The aim of this chapter was to detect the factors that might cause concrete massiveand buttress dams to fail. Especially to consider if other than the currently useddesign criteria could be relevant. Dams are designed against failure criteria basedon sliding and overturning. By going back and studying failures it is possible todetect if the failure criteria are sufficient or if other failure modes may have to beaccounted for in the design process.

This chapter includes a compilation of reported concrete dam failures across theworld; how and what have caused them to fail. This study excludes China since thedocumentation there is incomplete. Many failures occurred decades ago, and there-fore the documentation and important information regarding these failures mightbe inadequate.

Greater incidents of concrete dams were also included. Known dam failures withoutinformation about either the foundation or the failure cause were excluded.

4.1 Documentation of failures

The aim of the engineering industry today is to take responsibility of establishing aglobal collaboration as well as openness to share and increase the general knowledgeof the industry by creating formalised channels such as registers and organisations.The Committee on Dam Safety (CODS) in particular is working with this. However,it is difficult to obtain information about particular failures, especially in cases wherefailure took place long ago. Other reasons could be that some dam owners are notwilling to admit failure and do not make the records public, or due to a legal policypreventing publication of records. This has slowed down the technical developmentof the industry, limiting possibilities of understanding earlier generations’ thoughtsbehind their solutions and designs. The sharing of information and the ability totalk about dam failures could help increase our knowledge of the field as well asprovide the opportunity to learn from the experience of others, which would greatlyincrease the knowledge of the industry (Isander, 2013).

33

CHAPTER 4. FAILURE MODES OF CONCRETE DAMS

4.2 Compiled failures

The study includes failures of 19 concrete dams. These are divided into 12 massivedams, one gravity spillway dam and six buttress dams, presented in Table 4.1. Theincluded failures have, to varying degrees, documented information about the failureand the dam. For a full compilation of the studied dams, see appendix A.

Table 4.1: Concrete massive and buttress dams included in the study.

Dam name Country Dam typeHeight over

lowestfoundation

Year com-missioned

Year offailure

Bayless1 USA Massive 17 1909 1910(1911)Camara2 Brazil Massive 50 2002 2004Eigiau3 GB Massive 10 1911 1925Elwha river1(hydro-power)

USA Massive 51 1912 1912

High Falls6 USA Massive 9 1910 1999Marquette no 36 USA Massive 10 1924 2003Shih-Kang dam5

(gravity spillway)Taiwan Massive 22 1977 1999

St Francis1 USA Massive 62 1926 1928Torrejon-Tajo1 Spain Massive 62 1967 1965Upriver dam6 USA Massive 12 1937 1986Warrensburg6 USA Massive 8 1909 1976Xuriguera1 Spain Massive 42 1902 1944Zerbino4 Italy Massive 16 1925 1935Ashley1 USA Buttress 18 1908 1909Cascade lake dam8 USA Buttress 5 1908 1982Komoro1 Japan Buttress 16 1927 1928Morris Sheppard7 USA Buttress 58 1941 1986Overholser1 USA Buttress 17 1920 1923Stony creek1 USA Buttress 21 1913 1914

1(Douglas, 2002)2(Shaffner and Scott, 2013)3(J Andrew et al., 2011)4(Luino et al., 2014)5(Kung et al., 2001)6(Reegan, 2015)7(Anderson et al., 1998)8(Jarrett, 1986)

34

4.2. COMPILED FAILURES

4.2.1 Comparison of properties

In Section 4.2, Table 4.1 show the variation in year commissioned, age at failure andheight, these properties are compared in Figure 4.1 and Figure 4.2. The majority ofthe studied dams were commissioned before 1940 according to Figure 4.1.

0

1

2

3

4

5

6

7

Massive dams

Buttress dams

Figure 4.1: Year studied dams were commissioned.

0 2 4 6 8 10

During construction

During first filling *

During first five years

After five years

Buttress dams

Massive dams

Figure 4.2: Variation in age at failure of the analysed buttress and massive dams.

The buttress dams had according to Figure 4.2 slightly a higher tendency to failduring the first five years while the massive dams generally failed after five years.The majority of the failures did however occur within the first years. Failure is lessfeasible for older dams, where the possibility of failures per year decrease with theage of the dam.∗First filling is the first time the dam was filled

35


4.2.2 Failure type

Information about foundation material and geology of the foundation is presentedin Table 4.2. In some cases information is missing, and this is denoted with thesymbol "-". The failure of each dam is referred to a certain failure type, the usedfailure codes are defined below.

Ff, failure due to dam foundation.Fb, failure due to the structural behaviour of the dam body.Fa, failure due to appurtenant works.Fm, failure due to dam materials.

Table 4.2: Failure types for massive dams.

Dam name Foundationmaterial

Geology Failure code

Bayless Rock Sandstone horizontal layers with shale andclay between

Ff

Camara Rock Plane of micaceous silty clay FfEigiau Clay Hard blue clay containing boulders of granite

overlain by a layer of peatFf/Fm

Elwha river Soil/rock Fluvioglacial and conglomerate FfHigh Falls Rock - FmMarquette no 3 Rock - FaShih-Kang dam Rock Top deposition layer: unconsolidated gravel,

sands, silts and clay. On Soft bedrock:slate-gray, sandy-shale and silty-sandstones

Ffb

St francis Rock Conglomerate and schist FfTorrejon-Tajo - - Fa/FmUpriver dam Soil - FaWarrensburg - - FaXuriguera Rock - FfZerbino Rock Schist and hornfeld FafAshley Soil Fluvioglacial FfCascade lake dam Soil Glacial terminal-moraine sediments FfaKomoro Rock Tuff FfMorris Sheppard Rock Shale FfOverholser Rock - FfaStony creek Soil - Ff

36

4.2. COMPILED FAILURES

4.2.3 Failure mode

Table 4.3 contain information about the failure type by either information about thefailure mode of the foundation marked with the symbol "x", or the failure mode ofthe dam, as listed below. The failure mode is the parameter affecting the incidentmode. The following definitions have been used:

P, piping; water and material passing through the foundation.SC, scour; when sediment surrounding the dam abutment is removed.S, sliding.SH, shear sliding within dam.EQ, earthquake damage.T/C, tensile and compressive failure within dam.ST, structural damage to appurtenant equipment, such as spillway gates.

Table 4.3: Failure mode for massive dams.

Dam name Failure modefoundation

Failuremodedam

Failuretypecode

Failed due to overtopping

(P) (SC) (S)

Bayless x Ff Overtopping due to unknown causeCamara x Ff Not at highest water levelEigiau x Ff/Fm Not at highest water levelElwha river x Ff FilledHigh Falls (ST) Fm Overtopping due to unknown causeMarquette no 3 Fa Overtopping due to unknown causeShih-Kang dam (EQ) Ffb No informationSt Francis x x Ff Gradual during first fillTorrejon-Tajo (SH) Fa/Fm Flood during constructionUpriver dam x Fa Overtopping due to unknown causeWarrensburg (T/C) Fa No informationXuriguera x Ff No informationZerbino x x Faf Overtopping due to unknown causeAshley x Ff J ust spilling when pipe failedCascade lakedam

x Ffa Overtopping due to unknown cause

Komoro x x Ff No suggestions of high water levelMorrisSheppard

x Ff Releases kept within channel capacity

Overholser x Ffa Overtopping due to unknown causeStony creek x Ff Not clear if failed at top level or

overtopping

37


4.3 Description of failures

Descriptions about the failures for those dams which are not further discussed inSection 4.3.1 are given in Table 4.4 and Table 4.5.

Table 4.4: Failure description for massive dams.

Dam name Failure description

High Falls Overtopping led to breach of 23 meter long portionof concrete crest cap, left half the spillway. Repairscompleted.

Marquette no 3 Overtopping and failure of abutment, due to anfailure of upstream dam.

Torrejon-Tajo Shear sliding within the dam. Failure cause wastraced to organic material present in the aggregateand filling of the dam by a flood duringconstruction before the concrete had fully hardened.

Upriver dam Washout of the abutment and the power canalembankments due to overtopping. Not a completefailure and reparation of the dam was possible.

Warrensburg Breach of north abutment. Reconstructed in 1998.Xuriguera Failed by foundation sliding, shear strength and

poor design.

Table 4.5: Failure description for buttress dams.

Dam name Failure description

Ashley Piping failure in fine sand with clay and gravel, 6mdeep below cut-off.

Stony creek Piping in foundation followed by settling of dam,cracking and collapse of dam.

Cascade lake dam The dam was overtopped before tipping over andfailing. The cause of failure was the hydrostaticwater pressure on the dam and erosion of theabutments. Stored water was released rapidly dueto short time of breach development and the widthof the breach was large.

Komoro Failure due to softening of volcanic ash infoundation. Unclear cause, either piping, sliding orboth.

Overholser Overtopping leading to scour of abutment.

38

4.3. DESCRIPTION OF FAILURES

4.3.1 Documentation regarding failures

Bayless Dam, USA

Flaws in the Design of the Dam

A cut-off wall (shear key) was installed to suitable bedrock and steel rods were builtinto the wall and secured in the rock (The Engineering News Publishing company,1910). Against the upstream face, an embankment dam was placed, composed ofcompacted disintegrated shale, clay and some loam. The intention of the embank-ment was to prevent water percolating down to the porous strata beneath the dam.The engineer desired a deep cut-off wall that was to be constructed down throughthe rock strata, but was overruled due to its cost.

When the dam was completed one small vertical crack followed by more cracks,appeared on one side of the spillway. This was due to contraction since there wasno water in the dam. The Bayless dam prior to failure is shown in Figure 4.3.

Figure 4.3: Bayless Dam prior to failure (The Engineering News Publishing com-pany, 1910).

First failure

One year after the dam was completed, rapid melting of large amount of snowoccurred and within three days, the dam was filled to maximum capacity. The nextday, a large slice of earth below the dam dropped down 1.2 m and partially slid intothe valley. The water retained within the dam eroded a path beneath where theearth had slid. Eventually the water began flowing up through the ground in largequantities, 5-15 m downstream from the dam toe. The result was that the waterflowed under the dam in the embankment through the rock strata (The EngineeringNews Publishing company, 1910). During the third day, the flowing water resultedin that a portion of the dam, at the overflow spillway section, slided 0.4 m at thetop and 8 m at the base, causing the crack widths to increase at the downstreamface which unloaded the cracks at the upstream face. The movement lasted for eight

39


hours resulting in overflowing of the dam. It took 16 hours to completely emptythe reservoir. The total failure time of the dam was approximately two days. Thereservoir was lowered and no repairs were made before the dam was put back intoservice (DOI, 2012).

The initial cause of failure was piping, which caused softening of the clay and shale,lying between two layers of rock, causing the top layer of rock to slip forward onto thelower layer. The results from the failure were leaks under the dam, transverse cracksin the main section and movement of the central part downstream (The EngineeringNews Publishing company, 1910).

Second failure

At the end of summer, in 1911, the dam failed for the second time, due to highheadwater level, nearly as great as what caused the first slip failure. The dam failedin 30 minutes with no indications of a gradual failure. Four-fifths of the length of thedam broke into several large fragments, see Figure 4.4; most of them remained nearlyvertical. The two largest fragments near the centre (spillway section), fragment E &D in Figure 4.4 shifted downstream and rotated slightly from its original alignmentand, on both sides of these, large gaps were formed. At the west end, 38 m ofthe dam was still intact, fragment G in Figure 4.4 and to the east, fragment B& C the entire dam was broken and displaced. The failure was so extensive thatno conclusion of the initial point of failure could be drawn. The appearance of thefailed dam, however, indicates a sliding failure. Observations from the wreckage maysuggest that the westerly gap with its sections, slit and sheared out at levels abovethe foundation, as a secondary effect (The Engineering News Publishing company,1911).

Figure 4.4: Bayless Dam, resulting fragments and positions after failure (The Engi-neering News Publishing company, 1911).

40


Camara Dam, Brazil


The dam failed due to lack of information regarding the foundation. The smoothfoliation surfaces were left untreated leading to failure during the first filling. Indica-tions from material in the drains, leakage through the concrete and clogging of drainholes, etc. suggests that failure was about to occur. The designer recommended emp-tying the reservoir however due to political reasons it was not implemented (Shaffnerand Scott, 2013).

Figure 4.5: Camara Dam, failure at foundation of left abutment (Risk AssessmentInternational, 2013).

Failure

The reservoir was filled rapidly in about two weeks, due to heavy rains, and theretained volume gradually increased thereafter. After about five months when thereservoir was about 3/4 full, it failed at the foundation of the left abutment, dueto erosion from soil-filled discontinues. High pressure gradients developed underthe dam. As the flow rate of the reservoirs’ retained water increased the erosionand driving forces on the low-shear strength rock slabs, the dam and the abutmentstarted to slide until a hole was piped beneath the dam causing a flood. The remainsare shown in Figure 4.5. The remaining rock in the left abutment is essentially freefrom fractures indicating that piping occurred along the entire length of fracture(Ferc Engineering Guidelines, 2014).

41


Eigiau Dam, United Kingdom


Different contractors were involved in the construction of the dam. The footingsshould have been founded 1.8 m below the clay surface, but at the point of failure,only 0.5 m was embedded into the clay layer. From investigations, it was suggestedthat poor quality concrete contributed to the failure. The concrete lacked the correctvolumes of sand and cement, resulting in the aggregates not cementing together. Thestone aggregates were larger than desirable and were placed carelessly; in severalcases with voids under their bed surfaces (J Andrew et al., 2011).

Figure 4.6: Eigiau Dam, remaining breach (Geograph, 2010).

Failure

A 10 m long breach occurred in the concrete at the side leg of the dam, see Figure4.6. The breach scoured a 20 m wide channel three meters below the ground surface.Large portions, 1.5 million cubic meters, of water were released in the first hourand caused Coedty dam 2.5 miles downstream to overtop and the concrete wall tocollapse. In itself, the soft, porous condition of the boulder clay was aggravatedby cracking due to dehydration in the preceding summer when the lake bed wasexposed (J Andrew et al., 2011).

42


Elwha Dam, USA


The dam, seen in Figure 4.7, was founded on a deep gravel deposit and water there-fore blew out the foundation (Oakes, 2001). The contractors ignored specificationsfrom the engineers, resulting in the dam not being secured to the bedrock.

Figure 4.7: Elwha Dam prior to failure (KPLU 88.5, 2011).

Failure

The dam failed during the first fill due to piping failure of the alluvium under thedam. Lower sections of the dam were removed by the flowing water and largeportions of material eroded under the dam in about two hours resulting in a hole.

Various reparation methods were attempted and finally the hole was filled with de-bris. The Elwha dam failure started serious environmental investigations, regardingremoval of the dam to restore the river to its natural self-regulating state. The damwas finally removed 2012.

43


Morris Sheppard Dam, USA


The stability of the foundation was based on peak shear strengths and did notconsider uplift pressures (Anderson et al., 1998).

Failure

Sliding of the spillway section, see Figure 4.8, on the shale foundation, most probablyoccurred over a period of several years and was discovered during a routine inspection45 years after construction.

Figure 4.8: Morris Sheppard dam prior to failure, spillway section (Anderson et al.,1998).

Floatation of the lower portion of the hollow spillway section, together with thelow resistance to sliding, added to the tendency, of the hollow spillway to movedownstream. The reservoirs’ water level was quickly lowered to avoid completefailure.

The metal survey points, that were installed along a line, had formed a bow, whichindicated that the hollow spillway section had moved downstream on a slippage zonein the foundation, and from observations, cracks in the footings were found.

Core borings were made, which indicated that the hydrostatic uplift pressure, underthe spillway slab, was 65 % of the retention level. From the borings a longitudinalcrack along the top of the upstream cut-off was located. This may have allowed forwater to enter the foundation causing significant pressure beneath the shale layer.

Drainage wells were installed to reduce the uplift pressures. A network of measur-ing equipment was installed as to keep track of any movement or change in upliftpressure.

44


Shih-Kang Dam, Taiwan


The Shih-Kang dam was designed according to the traditional design concept of apseudo static earthquake acceleration (Kung et al., 2001). The effect from the ver-tical motion was, however, neglected. The original pseudo static horizontal acceler-ation was less than the real peak horizontal acceleration of the Chi-Chi earthquake,which, in turn, caused sliding failure.

Failure

The dam was damaged by surface ruptures caused by; an active fault, the largedisplacement of ground surface and great ground motion induced by the Chi-Chiearthquake. The dam moved in a north-west direction, 10 m vertically and 11 mhorizontally. The ground deformation caused the dam body to crack and separatefrom the foundation. The stiffness of the structures affected the deformation of thedam body.

The left side rock gradually rose towards the upper end of the fault-created scarp.The north part, between the spillway and the abutment, was cracked along the con-struction joints into several huge blocks, see Figure 4.9. Water then leaked throughthe cracks. The entire dam body seemed to have bent towards the downstream side.

Figure 4.9: Shih-Kang Dam post failure (HSS, 2013).

The water level was decreased to reduce pressure on the dam and complete failurewas avoided.

45


St. Francis Dam, USA


The dam, see Figure 4.10 was not designed with the correct uplift theory; upliftpressure acting to destabilise the sloping abutments. The cross section containedlimited seepage relief, only a few uplift relief wells beneath the central core and nooverlapping for expansion joints. To increase reservoir storage the dam height wasincreased 6 m without any substantive widening of the dam base width. The dam wasarched upstream but arch action was neglected in the design. During initial fillingseveral transverse cracks appeared which were filled in and sealed. The engineersdid not understand the concepts of effective stress and uplift. If the foundations hadbeen deeper, a cut-off wall and a grout curtain, with seepage relief wells would havebeen installed, the failure may had been avoided (Rogers, 1995).

Figure 4.10: St:Francis Dam prior to failure (Water and power, 2015).

Failure

The reservoir had been held within 7 mm from the spillway for five days and the dambecame unstable. At the west abutment a new larger leak and soiled discharge weredetected on the morning of the failure, caused by hydraulic piping. The dams’ westabutment, built up on a fault contact, was unknowingly founded upon massive paleomega-slides, causing cracking between the rock in the mid valley and the abutment(Rogers, 1995). Eventually the entire left hand side failed, inducing a domino effectof block failure at the right abutment (Veale and Davison, 2011). The blocks cutoff along the transverse crack resulted in large leaks at the toe of the east abutmentslide and 12 hours later, the dam collapsed. Forty minutes prior to the failure, thewater level was rapidly decreasing.

During final filling, a massive land slide occurred along the dams’ left abutment,carrying blocks and cutting off parts of the dam as the reservoir rose to full pool.The slide material initially plugged the outflow until the slide material was eroded by

46


the increasing flood wave. Full hydrostatic pressures entered the transverse crackscausing hydraulic uplift to eliminate the dams’ stabilising dead load. A sudden anddangerous overstressing of the dams’ cantilevered load capacity lead to excessive tiltand overturning, causing the upstream heel to go into tension, full hydrostatic headpressure was introduced within the dam structure which then finally cracked. Thecentral core of the dam was tilted towards the east abutment and tension developedin pre-existing cracks. Finally the west abutment was scoured away causing thecentral core to rotate (Rogers, 1995).

The dam appeared normal shortly before the failure; it failed 7.5 minutes later andemptied within an hour. All that was left was a single monolith standing in themiddle of the valley, see Figure 4.11.

Figure 4.11: St:Francis, resulting blocks from the abutment failure (Los AngelesTimes, 2013).

Zerbino Dam, Italy


The dam was constructed 300 m west of the Main Dam of Bric Zerbino, wherea saddle, formed by two ridges, was at a lower elevation which could have beenoverflowed and poured out into the riverbed (Luino et al., 2014). The dam, shownin Figure 4.12 was built rather hastily and without sufficient geologic investigations.It was assumed that the saddle consisted of sound rock. The dam stood on highlyjointed schist’s making up a particularly weak zone within the rock mass. Waterleaks were noticed across the rock diaphragm. Attempts to make the rock massimpervious were made with no satisfactory results. Miscalculation of precipitationin the area caused the water levels in the dam to differ from the values used in thedesign.

47


Figure 4.12: Zerbino Dam prior to failure (Molare.net, 2010).

Failure

Heavy rains caused the level of the reservoir to rise quickly and therefore the bottomdischarge valves were opened to handle these large volumes. The large amount ofmud and debris accumulating at the bottom of the dam caused the valves to stopworking after only a few minutes. The reservoir could then only be discharged at thesurface spillway and the siphons. Water then started to overflow, see figure 4.13 intoboth the main and the secondary dam, Zerbino. This caused repercussion on thejointed rocks that made up the saddle. After just one hour, the large water pressure,of the overflowing reservoir, displaced large rock blocks causing destabilisation of thedam which lead to the collapse. The Zerbino dam failed due to scour and slidingduring overtopping. The level of the reservoir went down by some 25 m within a fewminutes.

Figure 4.13: Zerbino Dam, flooded reservoir (Molare.net, 2010).

48

4.4. RESULTS OF THE COMPILED FAILURES

4.4 Results of the compiled failures

The comparison of the properties for the documented failures in Section 4.2.1 in-dicate that majority of the massive dams failed after five years, in fact in a widerange of 10-90 years after the dam was commissioned. This corresponds well withthe fact that the majority of the failed dams were built 1900-1940. The majorityof the dams stood for many years, which could indicate ageing of concrete as acontributing factor. All but one of these dams failed due to overtopping, showingthat the massive dams were sensitive to additional loading. The massive dams couldhave been designed against lower loads than they were subjected to at the time offailure. The majority of the buttress dams failed during the first five years, due tofoundation failure.

The failure types described in Section 4.2.2 are compared in Figure 4.14. The usedfailure codes are listed below.

Ff, failure due to dam foundation.Fb, failure due to the structural behaviour of the dam body.Fa, failure due to appurtenant works.Fm, failure due to dam materials.

0

1

2

3

4

5

6

Ff Ffa Ffb Ff/Fm Fa/Fm Fa Fm

Massive dams

Buttress dams

Figure 4.14: Failure types of the studied massive and buttress dams.

Of the studied failures, the majority failed due to, or partly due to, the dam founda-tion. This indicate the importance of a good foundation, meaning that informationabout the material and properties of the foundation play a crucial part in dam de-sign i.e. dam stability. This could be achieved by geological studies, core samples,material testing, site visits, etc.

49


Figure 4.14 indicates that the failure type for massive dams are mainly due todifferent combinations of foundation and dam failure. Massive dams are heavystructures with a large amount of concrete volume, which requires a strong enoughfoundation and correct casting arrangements. For buttress dams, foundation failureis the dominant failure mode. The reason could be that buttress dams are lightstructures. The casting sequence and construction of buttress dams are complicatedwhere faults are more prone to occur.

The failure modes descried in Section 4.2.3 are compared in Figure 4.15, with thedefinitions listed below.

P, piping failure.SC, scour failure.S, sliding failure.SH, shear sliding within dam.EQ, earthquake damage.T/C, tensile and compressive failure within dam.ST, structural damage to appurtenant such as spillway gates.No, no information was available.

0

1

2

3

4

P S/SC S S/P SH EQ Noinfo

T/C ST SC

Massive dams

Buttres dams

Figure 4.15: Failure modes of the studied massive and buttress dams.

The majority of the failures occurred in the foundation, seen in Figure 4.15, whichin some cases could have been avoided if improved geotechnical investigations wereperformed. Figure 4.15 also shows that failures in buttress dams are mainly causedby piping, sliding or both. Since not so many cases of failed buttress dams areincluded in this study, it is hard to conclude if there is a coincidence or if sliding andpiping is what mainly causes the buttress dams to fail. From the results, piping affectbuttress dams more than massive dams, not surprising considering that the water

50

4.4. RESULTS OF THE COMPILED FAILURES

only have to travel underneath the frontplate of the buttress. For a massive damthere is a much longer distance for the water to erode before it causes failure. Slidingis more common for buttress dams, also not surprising since the less use of concretevolume, and therefore also the vertical component acting on the sliding surface issmaller. This is not always the case for the buttress dams since the inclination ofthe frontplate results in an additional vertical water load.

The initial, or the main cause of failure is shown in Figure 4.15, although whenthe failure starts to propagate it will proceed to fail due to other failure modes orcombinations of failure modes. The dam failures described in Section 4.3.1 above,give a clear indication that there is a combination of failure modes leading to thefinal failure of the dam.

Out of all the studied cases, none failed due to global overturning; this indicates thatit is mostly a theoretical failure or that this failure mode is designed with high safetymargin. The only indications of overturning were found after the initial failure, as alocal failure of parts of the whole structure. This might create questions regardingthe importance of this design criterion, however it is still an adequate and accurateindicator of the dams’ stability.

As shown in Figure 4.15, only one dam failed within the dam body, which occurredduring construction before the concrete had cured properly, which therefore does notsuggest that the global stability is insufficient, but rather that special considerationsshould be made also during construction.

Figure 4.16 shows a summation of failures caused by faults in the dam design orduring construction of the dam.

0 5 10 15

Unknown

Poor construction

Inadequate groundinvestigations

Number of failures

Figure 4.16: Faults in dam design or construction for the studied dams.

From the failures described in Section 4.3.1 above it is clear that, to some extent,the neglect of vital information, or requirements, occurred in all cases. It is hard

51


to know if there were any relevant reasons for these assumptions, especially in thecases of poor construction.

Failure due to overtopping is shown in Figure 4.17, where it is detected that extremefloods affect the behaviour of the dam. Even though the height of the water levelis not the decisive factor for the failure, high water levels result in higher waterpressures affecting the concrete.

0 1 2 3 4 5 6

No suggestions of high waterlevel

Overtopping

Before first fill

Not at highest water level

No information

Buttress dams

Massive dams

Figure 4.17: Failure due to overtopping for the studied buttress and massive dams.

52

Chapter 5

Stability analyses

In the analytical analyses, several different dams were studied with varying geometryand loading conditions. Stability calculations were performed with Matlab R2013aaccording to both RIDAS and Eurocode as defined in Section 3.1 and Section 3.2.

Some of the analysed dams do not fulfil today’s failure criteria, due to that the damswere designed according to different presumptions. Improvements have in manycases been performed to fulfil the more recent criteria. The structural rehabilitationwere for some cases included, depending on the obtained drawings and information.

The dams were subjected to the first normal load case according to RIDAS, with theheadwater level of the dam, full ice load and all gates closed. For the calculationsaccording to Eurocode, the loads were combined according to design approach 3,see Section 3.2.1.

In addition the stability analysis tool, CADAM was used for the massive dams forcomparison. The failure criteria were calculated and compared to the analyticalcalculations.

The analytical calculations according to Eurocode were further investigated througha parametric study. The aim was to establish the most influential parameter, mod-ify the design loads by adjusting the partial factor and finally obtain results inagreement with the results from the stability calculations according to RIDAS.

The dams were also analysed analytically for the failure mode limit turning. Thedam with the largest difference in safety factor for overturning and limit turning, inaddition to a dam with a minimal difference in the safety factor for overturning andlimit turning, was analysed with the FEM software BRIGADE Plus 5.2.

5.1 Studied dams

The calculations were performed for a variation of dam types subjected to differentloads. The geometry of the different types of studied dams is shown in Figure 5.1and Figure 5.2. These figures show the headwater level, the rotation point and the

53

CHAPTER 5. STABILITY ANALYSES

characteristic loads applied to the different dam types, i.e. tailwater level and soillevel.

Figure 5.1: Geometry of the spillway Dam 9 (Left) and pillar Dam 11 (Right)

Figure 5.2: Geometry of the buttress Dam 14 (Left) and massive Dam 2 (Right).

5.1.1 Input data

The required information about the input data for the different dams was collectedfrom Section 3.1.1 for calculations according to RIDAS and Section 3.2.1 for calcu-lations according to Eurocode.

Recommended material values, characteristic values and partial factors were chosenaccording to Eurocode to the extent it was possible. Standard values from RIDASwere used if the required information was not found in Eurocode, which was thecase for both the uplift pressure and the ice load. Usually, the uplift pressure forpillars is calculated with a distribution for massive dams to account for uncertaintiesregarding the uplift pressure. This was applied to the analysed pillars in this report,instead of the distribution suggested in Section 3.1.1 according to RIDAS.

For the calculations performed according to RIDAS, the density of waterρ = 1000 kg/m3 and the gravitational force g = 9.81 m/s2 was used.

54

5.1. STUDIED DAMS

In this report dams were classified as safety class 3, since dam failures could resultin the loss of human life and are likely to have great economic consequences. Thepartial coefficient γd = 1.0 was applied to the loads according to Eurocode, see Table3.7.

Eurocode does not mention when rock bolts should be accounted for and thereforethe guidelines from RIDAS were applied. For calculations according to Eurocode,the common reinforcement steel Ks40 was used, with a characteristic strength fyk =370 MPa for φ = 25 mm and fyk = 350 MPa for φ = 32 mm (Ljungkrantz et al.,1994). For the calculations according to RIDAS, the load capacity of 140 MPa wasused for the rock bolts.

In the calculations according to Eurocode, the lateral earth pressure KO was calcu-lated according to Equation (3.7) from Section 3.2.1 with OCR = 1. The knowledgeof the shear resistance Td for the studied dams was limited and hence it was excludedfrom the calculations.

In Eurocode, the geotechnical design is more based on investigations compared toRIDAS. Since the value of the friction angle is not defined in Eurocode, a standardvalue of δd = 45◦ for rock foundations was used for the calculations of the slidingcriterion. For calculations according to RIDAS, values from Table 3.3 in Section3.1.2 was used for µmax to calculate the stability against sliding.

55


Geometry

Table 5.1 provides an overview of the geometry of the studied dams. In some casesthere is no value assigned to the geometric parameter, and this is denoted with thesymbol "-". Under the column for drainage "x" is placed if the dam has a drainagesystem. The following abbreviations are used in Table 5.1 and Table 5.2:

M, massive dam.B, buttress dam.S, spillway.P, pillar.S + P, spillway and pillar.GW, width of massive/pillar/spillway.

Table 5.1: Input data for the geometry used in the stability calculations.

Dam Type HeightGW Buttresswidth

Frontplatewidth

Inclinationupstreamface

Inclinationdownstreamface

Inclinationsliding plane

Drainage

[m] [m] [m] [m] [◦]

Dam 1 M 5 10 - - 1:1 10:6.5 0 -Dam 2 M 8 10 - - 1:1 10:6.5 0 -Dam 3 B 40 - 2 8 Varies 35:10 0 -Dam 4 B 12 - 2 8 45:10 14:10 0 -Dam 5 M 6 1 - - 10:1 2:1 11.3 -Dam 6 B 6 - 2 8 45:10 14:10 0 -Dam 7 M 13 1 - - 14:1 32:10 4.1 xDam 8 S 7 1 - - 1:1 Varies 4.8 -Dam 9 S 5 1 - - 1:1 Varies 0 -Dam 10 P 13 2.25 - - 1:1 Varies 16.4 -Dam 11 P 7 2.37 - - 1:1 8.5:1 0 -Dam 12 S 4 1 - - 1:1 Varies 0 -Dam 13 P 7 1 - - 1:1 35:10 0 -Dam 14 B 20 - 2 8 Varies 35:10 0 -Dam 15 P 16 4 - - 1:1 25:10 0 x(S+P) S 2 17 - - Varies Varies 0 xDam 16 P 19 3.75 - - 1:1 Varies 0 -Dam 17 P 19 2.65 - - 1:1 Varies 0 -Dam 18 P 18 1.85 - - 1:1 Varies 0 -

Loads

Table 5.2 provides an overview of the applied loads acting on the dams. When aload was not applied to a dam it is denoted with the symbol "-".

56

5.1. STUDIED DAMS

Table 5.2: Input data for the loads used in the stability calculations.

Dam Type IceLoad

Soilmaterial

Rockboltsφ

Strengthtendons

Headwaterlevel

Soil level Tailwaterlevel

Rotationpoint

[kN] [mm] [kN] [m.a.s.l]1 [m.a.s.l] [m.a.s.l] [m.a.s.l]

Dam 12 M 200 Moraine 25 - 745.21 744.9 - 740.5Dam 2 M 200 Moraine - - 745.21 744.9 - 738.1Dam 3 B 200 - - - 273 - - 235.4Dam 4 B 200 Rockfill 25 - 221.4 220 - 210.0Dam 5 M 100 Moraine - - 107.3 107.3 - 103.3Dam 6 B 200 - 25 - 221.4 - - 216.0Dam 7 M 100 Moraine - - 107.3 105.5 99.5 96.3Dam 8 S 25 - 32 - 107.3 - 100.6 98.4Dam 9 S - - - 500 107.3 - 99.5 97.8Dam 10 P 100 Rockfill - 500 107.3 98.4 102.2 99.0Dam 11 P 50 - - - 37.47 - - 31.06Dam 12 S - Gravel - - 26.45 - - 22.31Dam 133 P 50 - - - 15.05 - - 10.75Dam 14 B 200 - - - 273 - - 254.64Dam 15 P 200 - 25 - 125.3 - 114.3 111.8(S+P) S - - 25 - 115.5 114.3 111.8Dam 16 P 200 - - - 181 - - 164Dam 17 P 200 - - - 181 - - 163.5Dam 18 P 200 - - - 181 - - 164.5

1 meter above sea level.2 Rock bolts accounted for according to RIDAS.3 Additional loads from a gate, including the dead weight as well as the hydrostatic loadacting on the gate.

Limit turning

The safety factor for limit turning was calculated with the crushing resistanceRcr = 20 MPa for all the studied dams, from Table 3.9. From the results of the ana-lytical calculations, the dam with the largest difference in the safety factor for over-turning and limit turning and the dam with the small difference in the safety factorfor overturning and limit turning, was analysed with the FEM software BRIGADE.Additionally parametric analyses of the crushing resistance Rcr was performed toobtain the relationship between the crushing resistance and the safety factor forlimit turning for the dams analysed with BRIGADE.

57


5.1.2 Previously studied dams

In the report by Fouhy and Rios Bayona (2014), Dam 4, Dam 6 and Dam 15-18 in Table 5.1 and Table 5.2, was studied. Their study contains a probability-based analysis for evaluation of the stability of dams to account for the omissionof uncertainties. The aim of their study was to find a reliability index, β-targetvalue, applicable to the stability analyses for sliding and overturning of concretedams. Fouhy and Rios (2014) also performed a deterministic analysis to enable aninterpretation of the results from the probability-based analysis. They performedtheir analysis with three different load combinations, ’Load Combination 1’ whichcorresponds to the first normal load case in RIDAS was off interest for this report.

The differences compared to this report is presented below.

• The sliding criterion was based on the Mohr-Coulomb equation which includescohesion.

• Different values for concrete density.

• The ice load was based on a report by Adolfi and Eriksson (2013), which inturn is based on collected values for the annual maximum.

• A higher value for the characteristic yield strength of steel was used.

The results from Fouhy and Rios for load combination 1 are presented in Table 5.3.A brief comparison of the β-target values with the values stated in Eurocode, seeTable 5.4, was included in the report. The β-target values have a reference period ofone year, which correspond to the probability of dam failure evaluated over a periodof one year (Westberg, 2010).

The highlighted values in Table 5.3 are those who do not pass the criteria, eitheraccording to RIDAS for the deterministic calculations or according to Eurocode,RC3, for the probabilistic calculations. This enable us to easily note how highβ-values that were obtained and the difficulties in comparing these to Eurocode.There is a bad correlation between the β-value and the safety factor. For the slidingcriterion only one dam fail according to the deterministic calculations while all butone fail in the probabilistic calculations. For the overturning criterion all valuesobtained from the probabilistic calculations are well above the limit according toEurocode.

58

5.2. STABILITY CALCULATIONS

Table 5.3: Results from Fouhy and Rios Bayona (2014).

DamProbabilistic Deterministic

β for overturning β for sliding Overturning Sliding

Dam 4 8.57 3.93 1.56 0.57Dam 6 14.98 5.01 1.21 1.07Dam 15 8.54 2.12 1.57 0.65Dam 16 12.37 2.47 2.29 0.74Dam 17 11.53 5.24 2.1 0.56Dam 18 11.69 2.77 2.07 0.73

Table 5.4: Recommended minimum reliability values according to Eurocode 1990.

Safety class Minimum values for β Probability of failure/year

RC3 5.2 10−7

RC2 4.7 10−6

RC1 4.2 10−5

5.2 Stability Calculations

5.2.1 Design approaches

Overturning

The failure criterion for overturning was calculated according to Equation (3.1) fromSection 3.1.2. The safety factor s should satisfy Equation (5.1).

s =Mstab

Mover

> 1.5 (5.1)

In the calculations according to Eurocode, the failure criterion for overturningEquation (3.10) from Section 3.2.2, should satisfy Equation (5.2).

s =Md,stb

Md,dst

≥ 1.0 (5.2)

59


Sliding

The failure criterion for sliding according to RIDAS, was calculated according toEquation (3.2) from Section 3.1.2. The friction coefficient µ should satisfyEquation (5.3), with µmax for the rock foundation, from Table 3.3.

µ =RH

RV

≤ µmax = 0.75 (5.3)

The failure criterion for sliding Equation (3.11) from Section 3.2.2, according toEurocode, should satisfy Equation (5.4).

Hd −Rp;d

Rd

≤ 1.0 (5.4)

Limit turning

The safety factor was set to fulfil the value for overturning according to RIDAS,from Table 3.2, and should satisfy Equation (5.5).

The safety factor for limit turning, about the O axis:

Fs =ΣMr

ΣMt

> 1.5 (5.5)

Calculation method

The calculations were executed in Matlab where a numerical analysis tool wasdeveloped for the stability analyses. These analyses were performed according toFigure 5.3.

60

5.2. STABILITY CALCULATIONS

Stability

OVERTURNING RIDAS

SPILLWAY MASSIVE BUTTRESS

Define attributes Frontplate Hole Gate

Plot dam

Calculate geometry

Calculate Loads

Water pressure Horizontal

water pressure Vertical water

pressure if inclined upstream face

Vertical water pressure if spillway

Earth Pressure

Vertical and horizontal earth pressure: Soil same

density Soil diff.

density

Uplift

Horizontal and vertical uplift: If tailwater If hole If inclined

foundation else Vertical uplift

Rock bolts /tendons Horizontal

force Vertical

force

Ice load Horizontal force: If not

buttress else

Tail water pressure Vertical tailwater

pressure Horizontal

tailwater pressure if inclined downstream face

Input: Rotation points’ coordinates

Level arms

Loads Horizontal Vertical

Rotating moment Stabilising moment

Partial factors

Partial factorsLoad combinations

If Buttress

Input: Rotation points’ coordinates

Level arms

LoadsHorizontal Vertical

Partial factorsLoad combinations

SLIDING CRITERION

SLIDING CRITERION

RIDAS

OVERTURNING EUROCODE

If not Buttress

Input: Rcr

Position of O axis

Foundation crushing resistance

LIMIT TURNING

Partial factors

Rotating moment Stabilising moment

Input: Rcr Position of O

axis Foundation

crushing resistance

PILLAR

Figure 5.3: Stability calculations in Matlab.

61


5.2.2 Parametric study

In the calculations according to Eurocode, the safety is defined on the loads by apartial factor. Since these calculations were performed according to design approachDA 3 for retaining walls, the values for the partial factors may not be adequate forconcrete dams, therefore a parametric study was performed. The aim was to definethe parameter with the greatest influence on the stability by calculation of theweighting factor, see Equation (5.6). The design load could then be modified byvarying the partial factor to obtain a safety factor in agreement with the resultsfrom RIDAS.

The parametric study was performed for the loads acting on the dams, i.e. theparameters for the sliding and the overturning criterion were analysed. The impor-tance of each load was calculated in relation to the total loads affecting the structure,according to Equation (5.6).

αi =xi√

x2i + ...+ x2n(5.6)

where

xi is the value for the load studied.

α is the weighting factor.

Equation (5.7) should then be fulfilled for all the included parameters.

α2i + ...+ α2

n = 1 (5.7)

5.3 CADAM

The computer program CADAM is a tool used to analyse structural behaviour andsafety for concrete massive dams. The program can be used to perform 2D analysesof a single monolith, assuming a unit thickness of 1 meter (Leclerc et al., 2001).

CADAM was used to analyse the stability for the massive dams included in theanalytical calculations. The intention was to investigate if the program is a usabletool and equivalent to the analytical calculations. CADAM enables analyses of thecrack length, which was utilised to examine the contact between the dam body

62

5.3. CADAM

and foundation, known from the literature study in Chapter 4 to have a significantinfluence on failure.

5.3.1 Stability calculations

In CADAM the gravity method is used to perform the stress analyses to determinethe crack lengths and the compressive stresses. Along with the stability analyses toobtain the safety margin against sliding and the placement of the resultant for allforces acting on the structure. This makes it possible to evaluate stability againstsliding and overturning of the dam (Leclerc et al., 2003).

The method is based on rigid body equilibrium to determine the internal forcesacting upon the joints in the dam and the rock-concrete interface and beam theoryto determine the stresses (Leclerc et al., 2001).

In CADAM sliding is defined as:

SSF =(ΣV + U) · tanφ+ cAc

ΣH(5.8)

where

ΣV is the sum of vertical forces excluding uplift pressure.

U is the uplift pressure force resultant.

φ is the friction angle.

c is the cohesion.

Ac is the area under compression.

ΣH is the sum of horizontal forces.

Overturning is defined as:

OSF =ΣMS

ΣMO

(5.9)

where

ΣMS is the sum of the stabilising moments.

ΣMO is the sum of the destabilising moments.

63


5.3.2 Modelling

Input data

The massive dams presented in Section 5.1.1; Dam 1, Dam 2, Dam 5 and Dam 7were analysed.

The density of the concrete was set to 2300 kg/m3. The friction angle was definedas 45 ◦ and the cohesion was assumed to be zero for all dams. The loads presentedin Table 5.2 were applied to the dams.

Model definition

The geometry was defined as for the analytical calculations and the masses and ma-terials were defined. The geometry of Dam 2 and the applied loads; water pressure,ice load and resultants of the earth support fill are shown in Figure 5.4.

Dam 7 have a complicated geometry of the crest and was therefore simplified dueto limitations in CADAM, which only allows solid geometries with straight lines.

Figure 5.4: Section of Dam 2, showing geometry and loading in CADAM.

Load combinations

For dams with earth support fill the force resultants for earth pressure from theanalytical calculations were applied, using the user defined loads as point loads.The loads where combined according to normal load combination.

64

5.4. FE-ANALYSIS

For Dam 1 with rock bolts, the resultant force of the bolts was applied as two pointloads and positioned where the rock bolts intersect the interface of concrete and therock. The force was applied with an elevation of 0.1 m in order for the program toapply the force on the structure.

Cracking and uplift options

Specification of the cracking option was defined as tensile strengths for crack initia-tion and propagation. The calculations were performed with constant uplift pressureand with modified uplift pressure after cracking of the dam initiated.

The drainage system for Dam 7 was calculated with the option USACE 1995, whichenables the reduction of the uplift pressure as stated in RIDAS. In addition theoption to reduce the effectiveness of the drainage after cracking beyond the drain,was used.

Output variables

CADAM may generate different output reports where the stability drawings wereof interest. The studied output variables were the safety factors for sliding andoverturning, calculated in CADAM according to Equation (5.8) and Equation (5.9),with and without modified uplift pressure. The program also provides informationabout the extent of cracking in the concrete and rock interface.

5.4 FE-analysis

A 2D linear elastic finite element analysis, was performed to evaluate if Fishmans’assumption that the failure mode limit turning is more likely to occur compared toan overturning failure. One additional analysis was performed, where the rock wasprovided with plastic properties to study the impact of limit turning. The analysiswas performed in BRIGADE.

To prevent that sliding occurred in the FE analysis, a high value for the frictioncoefficient was defined between the rock foundation and the concrete dam body.Thereby the monolith was forced to overturn. The destabilising loads were succes-sively increased in the analysis to capture the load capacity of the dam.

65


5.4.1 Studied dams

A massive monolith, Dam 2, was analysed since the overturning criterion was notfulfilled and the difference in the safety factor for overturning and limit turning wassmall, later shown in Section 6.2.1. The second studied monolith was the buttressmonolith, Dam 3, due to the relatively large difference in the safety factor for limitturning and overturning, later shown in Section 6.2.1. The magnitudes of the appliedloads were taken from the analytical analyses in Section 5.1.1, presented in Table5.2. The material properties are presented below.

Massive monolith, Dam 2

The concrete is of type K300 with similar properties to C25/30, used today. Theused values are presented in Table 5.5.

Table 5.5: Material properties of C25/30 (EC 2, 2011).

Density 2300 kg/m3

Young’s Modulus 31GPa

Poisson’s ratio 0.2

The rock foundation was assumed to be granite and the material properties arepresented in Table 5.6.

Table 5.6: Material properties of granite (Björnström et al., 2006).

Density 2300 kg/m3


Poisson’s ratio 0.2

The crushing resistance was set to Rcr = 20 MPa. Equation (3.13) was used toobtain the compressive stress at which the crushing zone would form.

Buttress monolith, Dam 3

The material values for the rock foundation are assumed equal to those presentedin Table 5.6. The concrete strength is assumed to correspond to C20/25 as shownin Table 5.7.

66

5.4. FE-ANALYSIS

Table 5.7: Material properties of C20/25 (EC 2, 2011).

Density 2300 kg/m3


Poission’s ratio 0.2

For the buttress monolith, the crushing resistance was set to Rcr = 20 MPa. Theyield stress of the rock was defined as 13.6 MPa according to Equation (3.13) andwas assigned to plot a graph showing the stress and strain relationship.

5.4.2 Model definition

Model

The geometry of the monoliths and the foundations was defined by creating parts,and modelled as solid 2D deformable bodies.

Plane stress elements with different thickness was used to enable the parts to havedifferent widths. For the massive monolith, the parts were assigned a thickness of1 m. For the buttress monolith, the foundation as well as the monolith was dividedinto two parts. The foundation part connected to the frontplate was defined withthe same width as the frontplate. The foundation part connected to the buttresswas defined with the same width as the buttress.

Interaction

For both monoliths the interaction between the surfaces of the dam body and thefoundation was achieved using a frictional contact definition. The normal behaviourwas described by a penalty formulation to allow for elastic slip. The penalty formula-tion approximates “hard” interaction, hard pressure-overclosure (Dassault Systèmes,2007). The friction coefficient was chosen to 10 to disable sliding. The reason forthis is that that otherwise the monolith would fail due to sliding.

For the buttress monolith, the surface connection between the two foundation partsand the interaction between the frontplate and the buttress was defined with tieconstraints.

67


Load procedure

The analysis was defined in different load steps. In the initial step, the initial inter-actions regarding contact behaviour between the different parts and the boundarycondition was applied. The boundary condition for the foundation was set to con-strain the bottom of the foundation. The dead weight of the materials was appliedin the second step to allow the normal force and friction force to stabilise. Thegravity loads were defined with a uniform acceleration in the vertical direction. Forthe massive monolith, earth pressure was applied to the monolith in an additionalstep before the design loads were applied.

To simulate the failure of the monoliths, the method of overload was used where thedesign loads were applied in the first stage and in the second stage the destabilisingloads were increased until failure was reached. The horizontal water pressure and theuplift pressure were increased by increasing the density of the water. The methodresulted in that the monolith was subjected to an increased load while the lever armdid not change (Nordström et al., 2015).

For the massive monolith, the destabilising loads were increased by applying ampli-tude to the loads in the same step as the design loads were applied. For the buttressmonolith, the loads were increased by applying additional destabilising loads in aseparate step after the design loads had been applied, shown in Figure 5.5. Thedesign loads and constrains applied to the massive monolith are seen in Figure 5.6.The hydrostatic loads and earth pressure were defined as pressure loads and the iceload was defined as a pressure load but with a uniform distribution.

68

5.4. FE-ANALYSIS

Figure 5.5: Design loads applied to the buttress monolith (Left) and the increaseddestabilising loads (Right).

Figure 5.6: Design loads applied to the massive monolith.

Mesh

The buttress monolith was assigned a free mesh, built up out of pre-defined meshpatterns, due to the complex geometries, see Figure 5.7. The foundations and themassive monolith, with a simpler geometry, were assigned a structural mesh seeFigure 5.7. Plane stress and plane strain elements were chosen where each nodehas two degrees of freedom. The strain components perpendicular to the elementfaces are zero and the loading acts in the plane of the element (Patzák, 2014).

69


Plane stress elements of type CPS4, were used for the buttress monolith where thethickness of the elements is small in relation to the width. An convergence test wasperformed, resulting in a mesh size of 0.5 m for the whole buttress monolith model.The number of degrees of freedom was 9720. Plain strain elements were used forthe massive monolith where the thickness is equal to unity, by using the elementtype CPE4R. A mesh size of 0.1 m was used for the whole massive monolith model,where convergence test was performed. The number of degrees of freedom was 7040.

Figure 5.7: Mesh of the massive monolith (Left) and the buttress monolith (Right).

70

Chapter 6

Results and discussion

In this chapter the results from the analytical calculations are presented. A com-parison of the stability calculations between RIDAS (2011), Eurocode and CADAMis performed. The results from the parametric study for the stability calculationsaccording to Eurocode are presented. The results from the analytical calculationsand the FE-analysis of limit turning are presented.

6.1 Analytical analyses

6.1.1 Design approaches

A compilation of the analytical results is presented in this section.

Overturning

According to RIDAS, the safety factor s should satisfy Equation (5.1), fromSection 5.2.1 where

s > 1.5

According to Eurocode, stability against overturning should satisfy Equation (5.2),from Section 5.2.1 where

Md,stb

Md,dst

≥ 1.0

The safety factor for overturning according to CADAM, Equation (5.9) from Section5.3.1, should satisfy Equation (5.1) according to RIDAS.

71

CHAPTER 6. RESULTS AND DISCUSSION

The number of analysed dams that satisfied stability against overturning accordingto RIDAS, Eurocode and CADAM are shown in Figure 6.1.

0

1

2

3

4

5

6

7

Massive Buttress Spillway Pillar Spillway+ Pillar

Total monoliths

Overturning RIDAS

Overturning Eurocode

Overturning CADAM

Overturning modifieduplift CADAM

Figure 6.1: Total dams that satisfy the failure criterion for overturning according tothe different methods.

As seen in Figure 6.1, there are four dams that satisfied the overturning criterionaccording to Eurocode but could not fulfil the criterion according to RIDAS. Thepillars represent the biggest difference in meeting the criterion according to Eurocodecompared to RIDAS.

According to the analysis performed in CADAM there is no difference in the numberof massive dams that satisfy the failure criterion.

The essential part of the results was if RIDAS and Eurocode account for stabil-ity against overturning with comparable safety factors. However, the majority ofthe dams gave the same results for the two criteria, the numerical values accord-ing to Eurocode always overestimate the safety of the dams compared to RIDAS.The results gave a clear indication that a parameter study was of interest for theoverturning criterion.

A compilation of how well the results from RIDAS compare with the results ac-cording to Eurocode is presented in Figure 6.2. This was done by normalising thesafety factors obtained from the two methods. It is clearly shown that the resultsfrom Eurocode overestimate the safety and that the safety factor most comparableto RIDAS still differ more than 10 %.

72

6.1. ANALYTICAL ANALYSES

0.4 0.6 0.8 1 1.2 1.4 1.6 1.80.4

0.6

0.8

1

1.2

1.4

1.6

SEC/SEC.l im

SRIDAS/SRIDAS.lim

± 10%

RIDAS−EC

Figure 6.2: The relationship between safety factors for overturning calculated ac-cording RIDAS and Eurocode, presented with normalised safety factors.

The obtained numerical values for the safety factors are presented in Appendix B,Table B.1 and CADAM in Table 6.1.

Sliding

To achieve stability against sliding according to RIDAS, the friction coefficient µshould satisfy Equation (5.3), from Section 5.2.1:

µ ≤ µmax = 0.75

Stability against sliding, according to Eurocode, is fulfilled if Equation (5.4) fromSection 5.2.1 is satisfied, i.e.

Hd −Rp;d

Rd

≤ 1.0

The safety factor for sliding is defined according to Equation (5.8) from Section5.3.1 in CADAM. Equation (6.1) was used to enable the results from CADAM tobe comparable with the other analytical results.

s =1

SSF(6.1)

Figure 6.3 shows the number of analysed dams, that satisfy stability against slidingaccording to RIDAS, Eurocode and CADAM.

73


0

1

2

3

4

5

6

7

Massive Buttress Spillway Pillar Spillway +Pillar

Total monoliths

Sliding RIDAS

Sliding Eurocode

Sliding CADAM

Sliding modified upliftCADAM

Figure 6.3: Total dams that satisfy the failure criterion for sliding according to thedifferent methods.

Sliding according to CADAM resulted in no difference for unchanged uplift pressure.There was a difference when the uplift pressure was modified after cracking initiated,due to the increase in uplift pressure resulting in the dams more prone to fail.

From Figure 6.3 only one more dam failed according to Eurocode, this indicatesthat the sliding criterion from Eurocode is comparable to the sliding criterion fromRIDAS. The results do however indicate that the safety criterion is slightly harderto satisfy according to Eurocode.

A compilation of how well the results from RIDAS compare with the results accord-ing to Eurocode is presented in Figure 6.4. This was done by normalising the safetyfactors obtained from both RIDAS and Eurocode. The figure shows that there isa big difference in the correlation between the safety factors from Eurocode andRIDAS.

74


0.4 0.6 0.8 1 1.2 1.4 1.6 1.80.4

0.6

0.8

1

1.2

1.4

1.6

SEC/SEC.l im

SRIDAS/SRIDAS.lim

± 10%

RIDAS−EC

Figure 6.4: The relationship between safety factors for sliding calculated accordingRIDAS and Eurocode, presented with normalised safety factors.

Overturning and sliding

The analysed dams which satisfy stability against both overturning and sliding fail-ure are presented in Figure 6.5. The results show that there is no greater differencebetween the two methods, even though more dams satisfy the criterion accordingto Eurocode for overturning. The reason for the good compliance seen in Figure6.5 is that the failure criteria are more consistent for sliding according to the twomethods.

75


0

1

2

3

4

5

6

7

Massive Buttress Spillway Pillar Spillway+ Pillar

Total monoliths

RIDAS

Eurocode

CADAM

Modified uplift CADAM

Figure 6.5: Total dams that satisfy both failure criteria according to the differentmethods.

The results in Figure 6.5 motivate the performance of a parameter study to obtainfailure criteria applicable to dams.

CADAM

The safety factors from CADAM are presented in Table 6.1, showing the resultsfor the unchanged uplift pressure and the modified uplift pressure after cracking,including the analytical safety factors from Section 6.1.1 according to RIDAS.

Table 6.1: Results from analytical calculations and CADAM.

Dam Height [m] Analytical (RIDAS) CADAM CADAM-modified upliftSliding Overturning Sliding Overturning Sliding Overturning

Dam 1 5 0.67 0.83 0.68 0.85 0.95 0.78Dam 2 8 0.48 1.0 0.51 0.92 0.75 0.82Dam 5 6 0.36 1.03 0.35 1.07 0.48 0.94Dam 7 13 0.29 1.7 0.22 1.60 0.23 1.54

For overturning, the safety factors in Table 6.1 with modified uplift pressure aftercracking are a bit lower than the safety factors for the unchanged uplift pressureas well as the analytical safety factors. The results from the analyses gave a lowerstability with modified uplift pressure after cracking, which was expected. Sliding

76


resulted in higher safety factors for the modified uplift, i.e. increase the chance fora sliding failure to occur. The result was that only Dam 1 did not fulfil the safetycriterion and Dam 2 was just on the limit, compared to the unchanged uplift whereall the studied dams fulfilled the safety criterion.

The safety factors calculated in CADAM were comparable to the analytical calcu-lations according to RIDAS. Showing that CADAM is a suitable design tool formassive dams when analysing the stability.

In addition to calculating the safety factors, the crack length in the contact surfacebetween the rock and concrete was calculated. The analyses were performed toobtain additional information about the stability. The percentage of how much ofthe joint that cracked is presented in Table 6.2.

Table 6.2: Resulting crack length, presented as crack percentage of the contact sur-face from CADAM.

Dam Height [m] Crack percentage of the contact surface [%]CADAM CADAM - modified uplift

Dam 1 5 100 100Dam 2 8 100 100Dam 5 6 86.5 100Dam 7 13 7.5 10.6

For Dam 1, Dam 2 and Dam 5, that did not fulfil the failure criterion for overturningaccording to RIDAS, seen in Figure 6.1, the cracking in the concrete and rockinterface was significant. In most of the cases the crack length was 100 %, seenin Table 6.2, for both the unchanged and modified uplift. From the analyses ofthe crack length only Dam 7 resulted in percentages that are reasonable for a damthat has not failed. For the other dams the percentage of 100 % definitely serve asindications of insufficient stability.

Since these dams have not failed in reality, there is a possibility that the dams aresubjected to lower loads than the required loads in RIDAS. The stability criterionfor overturning is known to be difficult to fulfil for low dams, confirmed by theresults from the analyses in CADAM. There could also have been faults done in thesimplifications or that the program presents an inaccurate estimation.

77


Effect of rock bolts

Only Dam 1 can according to RIDAS account for rock bolts in the stability criteria.The steel strength accounted for in the two methods differ significantly. The differ-ence between the two methods was determined by performing additional stabilitycalculations for all the dams with rock bolts, presented in Table 5.2. The results didnot show any difference for sliding, it did not affect which dams that fulfilled thesliding criterion. A difference could be detected for the overturning criterion pre-sented in Figure 6.6. The safety factors for RIDAS and Eurocode were calculatedaccording to Section 3.1.1 and Section 3.2.1, with rock bolts included in the stabilitycalculations. It was easily detected that more dams satisfied the overturning criteriaaccording to Eurocode. Figure 6.6 shows how well the safety factors calculated withthe two methods compare. As previously stated it was clear that the calculationsaccording to Eurocode result in much higher safety factors.

0.4 0.6 0.8 1 1.2 1.4 1.6 1.80.4

0.6

0.8

1

1.2

1.4

1.6

SEC/SEC.l im

SRIDAS/SRIDAS.lim

± 10%

RIDAS−EC

Figure 6.6: The relationship between safety factors for overturning with rock boltscalculated according RIDAS and Eurocode, presented with normalisedsafety factor.

It was difficult to determine if the steel strength according to RIDAS is reducedmore than necessary or if Eurocode overestimates the strength. In this report thedesign criteria according to RIDAS were followed for the calculations including rockbolts and the corresponding calculations according to Eurocode were modified togive similar results.

78


6.1.2 Parametric study

By performing a parametric study it was possible to detect which partial factorto modify to gain results that coincided with RIDAS. The results are presented inFigure 6.7 and Figure 6.8.

Horizontal water pressure

Ice pressure

Uplift pressure

22%

28%

50%

Figure 6.7: The most influential destabilising parameters for overturning.

The destabilising parameters with the most influence for the overturning criterionwere the vertical uplift pressure (50 %), followed by the ice pressure (28 %) and thehorizontal water pressure (22 %).

Horizontal water pressure

Uplift pressure

Ice pressure12%

18%

70%

Figure 6.8: The most influential destabilising parameters for sliding.

The most influential destabilising load parameters for the sliding criterion weredetermined. The calculations resulted in that the horizontal water load had the

79


greatest influence (70 %), followed by the vertical uplift pressure (18 %) and the iceload (12 %).

Overturning

Due to four more dams satisfy the overturning criterion according to Eurocodecompared to RIDAS, one additional analytical calculation was performed, with newpartial factors to obtain results corresponding to RIDAS. Since the uplift is themost influential parameter for stability against overturning, the partial factor wasmodified. Due to uplift being a geotechnical action, the partial factor was changedto the value for an unfavourable variable load, γQ = 1.4. This resulted in that oneadditional dam failed, i.e. in agreement with RIDAS.

Further analytical calculations were then performed where γQ was increased to 1.5for the uplift pressure. While the partial factors for permanent favourable loads waschanged to γG = 0.9 due to that the stabilising loads had a great influence in theparameter study.

With these modifications of the partial factors for the overturning criterion, theresults from Eurocode were more comparable to RIDAS. The numerical values forthe calculations with the modifications are presented in Table 6.3.

80


Table 6.3: Result from parametric study of the overturning criterion according toEurocode and RIDAS.

Damnumber

Eurocode fromSection 6.1.1

Eurocodemodified

RIDAS fromSection 6.1.1

Dam 1 0.77 0.66 0.83Dam 2 0.78 0.65 1.00Dam 3 1.75 1.52 1.96Dam 4 1.34 1.13 1.65Dam 5 0.79 0.66 1.03Dam 6 0.71 0.61 0.94Dam 7 1.50 1.12 1.71Dam 8 1.07 0.79 1.17Dam 9 1.79 1.23 2.04Dam 10 1.69 1.31 2.08Dam 11 1.15 0.91 1.31Dam 12 1.63 1.19 1.75Dam 13 1.17 0.98 1.44Dam 14 1.35 1.17 1.58Dam 15 1.49 1.07 1.67Dam 16 1.23 0.98 1.43Dam 17 1.38 1.07 1.58Dam 18 1.32 1.04 1.55

In Figure 6.9 it is possible to see how the modified partial factors result in safetyfactors that correspond better to RIDAS. An improvement of the previous resultswas obtained and the majority of the results do not differ more than ± 10 % fromthe safety factor from RIDAS.

81


0.4 0.6 0.8 1 1.2 1.4 1.6 1.80.4

0.6

0.8

1

1.2

1.4

1.6

SEC/SEC.l im

SRIDAS/SRIDAS.lim

± 10%

RIDAS−EC modified

RIDAS−EC

Figure 6.9: The relationship between safety factors for overturning calculated ac-cording RIDAS, Eurocode and Eurocode with modified partial factors,presented with normalised safety factors.

Sliding

RIDAS and Eurocode were quite consistent for the sliding criterion, but since somedifference between the safety factors was obtained, an attempt to achieve resultsequivalent to RIDAS was made.

The horizontal water pressure was the most influential parameter. The partial factorfor the horizontal water pressure was modified to γQ = 1.0, which did not affect theresults for the stability criterion. If γQ for the horizontal water pressure was furtherdecreased or further modifications of the partial factors for the loads were made,it resulted in dams already in agreement with RIDAS to change, i.e. an unwantedresult. The results gave that the partial factor for the horizontal water load couldonly be modified to γQ = 1.0.

As shown in Figure 6.10 the modified partial factors give a slightly better corre-spondence to RIDAS. The result is however, still not satisfying and therefore morestudies need to be performed to find partial factors that would result in a acceptablecorrespondence to RIDAS.

82


0.4 0.6 0.8 1 1.2 1.4 1.6 1.80.4

0.6

0.8

1

1.2

1.4

1.6

SEC/SEC.l im

SRIDAS/SRIDAS.lim

± 10%

RIDAS−EC modified

RIDAS−EC

Figure 6.10: The relationship between safety factors for sliding calculated accord-ing RIDAS, Eurocode and Eurocode with modified partial factors, pre-sented with normalised safety factors.

Rock bolts

Including rock bolts gave no difference in the result regarding the sliding criterionand therefore only the partial factors for overturning are discussed in this section.

The modified partial factors used for the results presented above, γQ = 1.5 andγG = 0.9, were used to determine if the results from Eurocode and RIDAS wouldbe more similar. The outcome was not as desired and an additional calculation wasperformed where, in addition to the previous adjustments, the partial factor γs waschanged to 1.35. The adjustment of γs affected the strength of the rock bolts. Themodifications resulted in more comparable safety factors to RIDAS, shown in Figure6.11.

83


0.4 0.6 0.8 1 1.2 1.4 1.6 1.80.4

0.6

0.8

1

1.2

1.4

1.6

SEC/SEC.l im

SRIDAS/SRIDAS.lim

± 10%

RIDAS−EC modified

RIDAS−EC

Figure 6.11: The relationship between safety factors for overturning with rock boltscalculated according RIDAS, Eurocode and Eurocode with modifiedpartial factors, presented with normalised safety factors.

6.1.3 Previously studied monoliths

From the report by Fouhy and Rios Bayona (2014), the results was interpreted as abig scatter, where the performed comparison with the acceptable values according toEurocode was limited. From the probabilistic analyses in their report, no conclusionscould be drawn about if the failure criterion was fulfilled or not. Fouhy and RiosBayona (2014) showed that their probabilistic method gave the same β-value for adam that fulfil the deterministic failure criterion as for a dam that fails to fulfil thecriterion. By observation and comparison to Eurocode, the probabilistic methoddid not result in comparable safety limits. The use of partial factors in this reportshows better correspondence with RIDAS.

6.2 Analyses of limit turning

6.2.1 Analytical analysis

The safety factors from the analytical calculations for limit turning and overturningaccording to RIDAS, calculated with Rcr = 20 MPa, are presented in Table 6.4. Thetable also includes the difference between the safety factors for the two criteria.

84

6.2. ANALYSES OF LIMIT TURNING

Table 6.4: Results from analytical calculations of stability against limit turning andoverturning.

Dam Height [m] Limit turning Overturning fromSection 6.1.1

Difference[%]

Dam 1 5 0.829 0.831 -0.2Dam 2 8 1.000 1.004 -0.4Dam 3 40 1.738 1.955 -11.1Dam 4 12 1.619 1.653 -2.1Dam 5 6 1.029 1.033 -0.4Dam 6 6 0.931 0.942 -1.2Dam 7 13 1.704 1.707 -0.2Dam 8 7 1.172 1.175 -0.3Dam 9 5 2.141 2.043 +4.8Dam 10 13 1.971 2.082 -5.3Dam 11 7 1.307 1.314 -0.5Dam 12 4 1.749 1.751 -0.1Dam 13 7 1.435 1.440 -0.3Dam 14 20 1.496 1.576 -5.1Dam 15 16 1.667 1.667 0Dam 16 19 1.403 1.429 -1.8Dam 17 19 1.565 1.579 -0.9Dam 18 18 1.526 1.549 -1.5

The dam with the greatest difference between the safety factors was Dam 3 wherethe difference between the two criteria was 11 %. For the majority of the dams thedifference between the safety factors was insignificant. For all dams under 10 m theimpact of limit turning on the safety factor was negligible. For Dam 9 the safetyfactor for limit turning increased, showing that limit turning not always result inlower safety factors. This is presumed to be due to that the sum of the stabilisingforces are either increased more or decreased less than the sum of destabilising forces.A high contributed stabilising force from the rock might be an other reason why thesafety factor increased. From the assumption that limit turning should fulfil thesame criterion as overturning, only one dam, Dam 14, failed due to limit turningwhile fulfilling the criteria for overturning. By the comparison of Dam 5 and Dam 6as well as for Dam 4 and Dam 7, the difference between the safety factors is greaterfor buttress dams compared to massive dams of the same height, highlighted in theresults presented in Table 6.4.

The impact of the crushing resistance on the safety factor for limit turning can beseen in Figure 6.12 and Figure 6.13. For Dam 2 in Figure 6.12, the relationship

85


between the safety factor for limit turning and the crushing resistance show thatthe limit turning criterion is of importance for rock with poor quality, typicallya crushing resistance less than 5 MPa. However, for Dam 3, in Figure 6.13, thesafety factor for limit turning does not approach the value of the safety factor foroverturning in the same way as for Dam 2, therefore in this case the rock qualitybecomes more important. The possibility of a limit turning failure may be worth toconsider for high dams.

0 5 10 15 20 25 30 35 40 45

0.6

0.7

0.8

0.9

1

1.1

1.2

Safety

factor[-]

Crushing resistance of rock [MPa]

Limit turning

Overturning

Figure 6.12: The relationship between the safety factor for limit turning and thecrushing resistance for Dam 2.

0 5 10 15 20 25 30 35 40 45

1

1.5

2

2.5

3

3.5

Safety

factor[-]

Crushing resistance of rock [MPa]

Limit turning

Overturning

Figure 6.13: The relationship between the safety factor for limit turning and thecrushing resistance for Dam 3

A significant influence of the crushing resistance can clearly be seen which also showthe need for geotechnical investigations in order for limit turning design criteria to be

86


useful. In dam design today approximated material values are often used. The useof approximated values for the crushing resistance may lead to inadequate design.

6.2.2 FE-analysis

For both the massive and the buttress monolith, the ultimate loading for the desta-bilising loads was defined as two times the design load. The ultimate loading waschosen to ensure the model to reach failure.

Dividing the load at which the monolith goes to failure with the design load made itpossible to extract the safety factor to examine if it corresponded with the analyticalcalculations from RIDAS.

By analysing if the compressive stresses that corresponds to the crushing resistancewere reached, it was possible to determine if limit turning failure would occur beforeoverturning failure. Based on Equation (3.13) a compressive strength of 13.6 MPa

(corresponding to the crushing resistance 20 MPa) was used to detect failure.

For Dam 3, the compressive stresses corresponded to the crushing resistance whenthe ratio of total loads/design loads = 1.68. The parts of the rock underneath thetoe started to plasticise, i.e. formation of the crushing zone, indicating that limitturning would occur before overturning. In Figure 6.14, when the ratio of totalloads/design loads = 1.89, a greater part of the rock beneath the toe had reachedstresses equal to the crushing resistance and the crushing zone was easily detected.This gives that the analytical safety factor for limit turning 1.74 from Section 6.2.1appears reasonable.

By extracting the stresses distributed over one element in the rock crushing zone,plastic deformation of the rock could also be seen, shown in the stress-strain curve inFigure 6.15. Showing that the stresses in the rock lead to the formation of a crushingzone before the overturning failure occurred, at a safety factor corresponding to 1.94,i.e. similar to the analytical safety factor for overturning.

87


Figure 6.14: Compressive stress in Dam 3.

0

2

4

6

8

10

12

14

16

0 0.0005 0.001 0.0015 0.002 0.0025 0.003 0.0035

Stress [M

Pa]

Strain [‐]

Figure 6.15: Stress-strain relationship for Dam 3.

88


The analysis of Dam 2 did not display any signs of limit turning failure, which wasnot surprising considering the results from Section 6.2.1 where Dam 2 only had0.4 % lower safety factor for limit turning compared to overturning. The modelfailed when the safety factor was 1.09, which corresponds to the analytical safetyfactor for overturning. The rock did not reach the compressive stresses required forthe crushing zone to develop, as seen in Figure 6.16. This indicates that the dam ismore prone to fail according to overturning compared to limit turning.

Figure 6.16: Compressive stresses in Dam 2.

89

Chapter 7

Conclusions

The objective of this report was to examine if the design criteria for concrete damsused today are enough or if modifications are needed. The other objective was toanalyse if Eurocode was comparable to RIDAS in dam design.

7.1 Failure modes of concrete dams

The majority of the studied massive dams, built before 1940, stood for many yearsbefore failure. The dams were designed under different circumstances and for thedams that did not fail during the first five years, there must be a different expla-nation. The explanation could be the unknown parameters such as climate changeaffecting the loads acting on the dam. The difficulties in predicting amount of rainand increasing natural disasters is a big problem with design today, and even morea problem a few decades ago. Today there are possibilities of computer simulationsfor probabilistic variations of different loads affecting the dam, resulting in betterdesigned structures. However, as seen in Section 4.4, this has not prevented recentdam failures, which were in some cases caused by faults in the design. Along withcomputer simulations, the development of concrete recipes has been ongoing. Today,different measures help improve the concrete dam design and achieve better stabilitysuch as different types of cement, e.g. Portland cement and improved quality of thereinforcement, e.g. profiled reinforcement bars.

To continue to learn from the previous dam failures helps the development of damstructures along with predictions about the future and improved concrete recipes.This is also true for buttress dams where the majority of the dams failed duringthe first five years. Since buttress dams are relatively new types of structures, they

91

CHAPTER 7. CONCLUSIONS

are limited to lessons from only a few previous failures. Additionally the buildingtechniques and material knowledge have improved for these structures, preventingfailure.

From the analysed failures, one result is that the dam foundation has a crucial partin a safe dam design. Failure in the dam foundation were often caused by the lackof information about the foundation materials. Establishing tougher regulationsconcerning implementation of actual ground investigations could be a solution. InSweden today, many of the calculations are based on standard values of the rockfoundation material instead of actual foundation properties, which could lead to anoverestimated safety.

The failure is often initiated by one specific failure mode resulting in a combinationof failures, which shows that pure failure modes seldom occur. There were failuremodes not covered, indicating that the design criteria are not thorough. The resultsin Section 4.4, indicate that the failures where more information was collected,were either caused by a poor design or poor workmanship shown by the contractorwhere short-cuts and lack of understanding are displayed. One method to increasesafe construction design, already established in Sweden, is tougher regulations andplacing responsibility on the dam owners, which makes them prone to design safeconstructions.

The known failures were studied to determine whether the currently used designcriteria would suffice. In conclusion, it was determined from the information regard-ing the collective failures, that the design criteria is extensive enough. However, thechallenge lies in ensuring that the construction of the dam is correctly performed tofulfil today’s criteria.

The results are summarised as; the current design criteria are sufficient and saferdam designs are achieved if the following conditions are considered:

• A thorough geotechnical investigation is performed.

• Engineers with enough knowledge are involved in designing the dam.

• Increased focus on technical issues rather than economic aspects.

7.2 Analytical calculations

Limit turning, which is not a design criterion in Sweden today, produced lowersafety factors for the majority of the dams, indicating that it might need to be

92

7.3. DESIGN GUIDELINES

considered. The simulations in the software BRIGADE showed that limit turningwas important for the highest of the two dams. Increasing height of the dam, resultin greater influence of limit turning. A difference in the effect of limit turning onbuttress dams compared to massive dams were detected.

The results indicated that the quality of the rock had a great influence on thesafety factor. The difficulties with limit turning appeared to be that geotechnicalinvestigations of the rock quality are essential for the calculations to be sufficientas a design criterion. Approximated values for the crushing resistance could lead toinadequate design. Limit turning also requires more extensive calculations makingit less convenient to use. Nevertheless, for buttress dams and high dams, especiallyhigh buttress dams, limit turning is of interest.

7.3 Design guidelines

Comparing the methods of partial factors and probabilistic analyses, the method ofusing partial factors appears to be easier to adapt to the requirements defined byRIDAS.

From the comparison of RIDAS and Eurocode it was clear that modifications wereneeded for Eurocode to apply to concrete dams, especially for overturning where tomany dams passed the failure criterion. By modifications of the mentioned partialfactors, Eurocode appeared to give similar safety factors as RIDAS. The sliding cri-terion appeared to be more difficult to adjust to RIDAS, even though the differencebetween the two methods was not that significant. The partial factors are presentedin Table 7.1, where the modified partial factors are highlighted.

93

CHAPTER 7. CONCLUSIONS

Table 7.1: The resulting modified partial factors for the failure criteria according to Eu-rocode

Load EC EC-modified overturning EC-modified sliding

Dead weight 1.0 0.9 1.0Horizontal water pressure upstream side 1.1 1.1 1.0Horizontal water pressure downstream side 1.0 0.9 1.0Vertical water pressure 1.0 0.9 1.0Vertical uplift pressure 1.1 1.5 1.1Horizontal uplift pressure 1.1/1.01 1.1/0.9 1.1/1.0Ice load 1.5 1.5 1.5Rock anchors2 1.1 1.1 1.1Earth pressure 1.1/1.0 1.1/0.9 1.1/1.0

1 unfavourable/favourable2 the partial factor for the reinforcement was modified from 1.15 to 1.35

The stability calculations will be more complicated in Eurocode compared to RIDAS,since the direction of the force need to be considered as well as if the load is afavourable or unfavourable. An additional disadvantage of using partial factors isthat different partial factors would have to be used for sliding and overturning. Theengineer would be responsible for interpreting the different loads affecting the specificdam analysed. Resulting in more factors to consider, leaving room for mistakes.

7.4 Future studies

Further studies needs to be performed to justify the statement that limit turning isof interest for buttress dams and high dams. For the method of using partial factorsto work, further studies need to be performed, including if it is possible to accountfor the strength of the rock bolts according to Eurocode and which partial factorsthat could be reasonable to utilise.

94

Bibliography

Adolfi, E., Eriksson, J., 2013. Islastens inverkan på brottsannolikheten för glidningoch stjälpning av betongdammar. Tech. rep., KTH Royal Institute of Technology.

Ali, M. H., March 2012. Comparison of Design and Analysis of Concrete GravityDam. Natural Resources 03, 18–28.

Anderson, C., Mohorovic, C., Mogck, L., Cohen, B., Scott, G., 1998. Concrete DamsCase Histories of Failures and Nonfailures with Back Calculations. Tech. rep.,United States Department of the Interior, Bureau of Reclamation, Dam SafetyOffice.

Andersson, C., 2012. SVC-dagarna 2012. SVC - Svenskt vattenkraftcentrum, 2012-09-25.

Andersson, C., 2014. Summering av workshop om Eurokoder med tillämpning påbetongdammar. Workshop 25/11 2014 om Eurokoder med tillämpning på betong-dammar. Energiforsk(Elforsk), 2014-11-25.

Bergh, H., 2014. Hydraulic Engineering. KTH Civil and Architectural Engineering,Sweden.

Björnström, J., Ekström, T., Hassanzadeh, M., 2006. Spruckna betongdammar –Översikt och beräkningsmetoder. Elforsk report 06:29, Eneriforsk(Elforsk).

Bond, H., 2014. Dimensionering och kontroll av betongdammar enligt RIDAS 2012.Workshop 25/11 2014 om Eurokoder med tillämpning på betongdammar. Energi-forsk(Elforsk), 2014-11-25.

Dassault Systèmes, 2007. Abaqus Analysis User’s Manual. Dassault Systèmes Group.

DOI, 2009. 19. Risk Analysis for Concrete Buttress Dams. Bureau of reclamination,United States Departement of the Interior.

DOI, 2012. 20. Risk Analysis for Concrete Gravity Structures. Bureau of reclamina-tion, United States Departement of the Interior.

95

BIBLIOGRAPHY

Douglas, K. J., 2002. The Shear Strength of Rock Masses. Ph.D. thesis, The Uni-versity of New South Wales.

EC 0, 2002. Eurocode - Basis of structural design. EN 1990.

EC 1, 2013. Eurocode 1: Actions on strucutres - Part 1-1: General actions - Densi-ties, self-weight, imposed loads for buildings. EN 1991-1-1.

EC 2, 2011. Eurocode 2: Design of concrete structures - Part 1-1: General rules andrules for buildings. EN 1992-1-1.

EC 7, 2011. Eurocode 7: Geotechnical design - Part 1: General rules. EN 1997-1.

EKS 9, 2013. Boverkets föreskrifter om ändring i verkets föreskrifter och allmännaråd (2011:10) om tillämpning av europeiska konstruktionsstandarder (eurokoder).BFS 2013:10, Boverket.

Ferc Engineering Guidelines, 2002. Chapter III Gravity Dams. Ferc - Federal EnergyRegulatory Commission.

Ferc Engineering Guidelines, 2014. Risk-Informed Decision Making, Chapter R5.Ferc - Federal Energy Regulatory Commission.

Fishman, Y. A., 2007. Features of shear failure of brittle materials and concretestructures on rock foundations. International Journal of Rock Mechanics and Min-ing Sciences 45 (6), 976–992.

Fishman, Y. A., 2009. Stability of concrete retaining structures and their inter-face with rock foundations. International Journal of Rock Mechanics and MiningSciences 46 (6), 957–966.

Fouhy, D., Rios Bayona, F., 2014. Reliability-Based Analysis of Concrete Dams.Master Thesis, KTH Royal Institute of Technology.

Geograph, 2010. The breach in the dam wall, Liyn Eigiau reservoir. http://www.geograph.org.uk/photo/1963073, retrieved 2015-04-23.

Gustafsson, A., Johansson, F., Rytters, K., Stille, H. k., 2008. BetongdammarsGlidstabilitet - Förslag på nya riktlinjer. Elforsk report 8:59, Energiforsk(Elforsk).

HSS, 2013. Shih-Kang Dam Project in Taiwan. http://www.hss.dk/casestudies/taiwan.aspx, HSS Engineering warning system solutions, retrieved 2015-04-23.

ICOLD, 1995. Dam failures statistical analyses. Bullentin 99. Commission Interna-tionale des Grands Barrages.

96

http://www.geograph.org.uk/photo/1963073

http://www.geograph.org.uk/photo/1963073

http://www.hss.dk/casestudies/taiwan.aspx

http://www.hss.dk/casestudies/taiwan.aspx

BIBLIOGRAPHY

Isander, A. (Ed.), 2013. Dams: Accidents and Incidents What Can We Learn. Eu-ropean Club of ICOLD, International Comission on Large Dams.

J Andrew, C., Tedd, P., Warren, A., 2011. Lessons from historical dam incidents.

Jarrett, R. D., 1986. Hydrology, Geomorphology, and Dam-Break Modeling of theJuly 15 , 1982 Lawn Lake Dam and Cascade Lake Dam Failures. Tech. rep., UnitedStates Department of the Interior.

Johansson, F., 2005. Stability Analyses of Large Structures Founded on Rock. Tech.rep., Department of Civil and Architectural Engineering, KTH Royal Institute ofTechnology.

Kleivan, E., Kummeneje, G., Lyngra, A., 1994. Concrete in Hydropower structures.Norwegian Institute of Technology, Division of Hydraulic Engineering.

KPLU 88.5, 2011. Elwha River dam removal historic, but not explosive. http://www.kplu.org/post/elwha-river-dam-removal-historic-not-explosive,news for Seattle and the Northwest, retrieved 2015-04-23.

Krounis, A., 2013. Uncertainty in Sliding Stability Analyses of Existing ConcreteGravity Dams with Bonded Concrete-Rock Interfaces. Ph.D. thesis, Division ofSoil and Rock Mechanics, Royal Institute of Technology.

Kung, C.-S., Ni, W.-P., Chiang, Y.-J., 2001. Damage and Rehabilitation Work ofShih-Kang Dam. Tech. rep., Sinotech Engineering Consultants, LTD.

Leclerc, M., Léger, P., Tinawi, R., 2001. Cadam User’s Manual. Montréal, version1.4.3.

Leclerc, M., Léger, P., Tinawi, R., 2003. Computer aided stability analysis of gravitydams - CADAM. Advances in Engineering Software 34, 403–420.

Ljungkrantz, C., Möller, G., Petersons, N., 1994. Betonghandbok - Material, 2ndEdition. AB Svensk Byggtjänst och Cementa AB.

Los Angeles Times, 2013. St. Fran Dam collapse. http://framework.latimes.com/2013/03/12/st-francis-dam-collapse, retrieved 2015-05-02.

Luino, F., Tosatti, G., Bonaria, V., 2014. Dam failures in the 20th century: nearly1,000 avoidable victims in Italy alone. Journal of Environmental Science and En-gineering 3, 19–31.

97

http://www.kplu.org/post/elwha-river-dam-removal-historic-not-explosive

http://www.kplu.org/post/elwha-river-dam-removal-historic-not-explosive

http://framework.latimes.com/2013/03/12/st-francis-dam-collapse

http://framework.latimes.com/2013/03/12/st-francis-dam-collapse

BIBLIOGRAPHY

Malm, R. (Ed.), 2015. Hydropower- Technology, Economy, Sustainability. KTH Con-crete Strucutures / SWECO, course material.

MATLAB, 2013. Matlab r2013a. The MathWorks, Inc.

Molare.net, 2010. The disaster of molare. http://www.molare.net/disaster/the_disaster_4.html, retrieved 2015-04-23.

Nilsen, B., Thidemann, A., 1993. Rock Engineering. Norwegian Institute of Tech-nology, Division of Hydraulic Engineering.

Norconsult, 2012. Lima kraftverk. https://www.norconsult.se/referenser/

2014/lima-kraftverk/, retrieved 2015-03-03.

Nordström, E., Malm, R., Johansson, F., Ligier, P.-L., Lier, Ø., 2015. Betong-dammars brottförlopp - Litteraturstudie och utvecklingspotential. Elforsk report2015:122, Energiforsk(Elforsk).

Oakes, R., 2001. Historical Background on the Elwha River Dams. American FieldGuide, Lakeridge High School.

Patzák, B., 2014. OOFEM Element Library Manual. http://www.oofem.org/

resources/doc/elementlibmanual/html/elementlibmanual.html, retrieved2015-04-17.

PTI, 2000. What Is Post-Tensioning. Post Tensioning Institute.

Reegan, P., 2015. Dam Incident Data. Ferc - Federal Energy Regulatory Commission,USA.

RIDAS, 2011. Kraftföretagens riktlinjer för dammsäkerhet, RIDAS. Avsnitt 7.3 Be-tongdammar Tillämpningsvägledning. Svensk Energi - Swedenenergy AB.

Risk Assessment International, 2013. Dam failure news. http://www.

risk-assessment.at/news_en.html, retrieved 2015-04-23.

Rogers, D. J., 1995. A man, a dam and a disaster: Mulholland and the St. Fancisdam. In: Doyance B. Nunis, J. (Ed.), The St. Francis Dam Disaster. HistoricalSociety of Southern California, pp. 1–109.

Sanscot Technology, 2015. Brigade Plus 5.2. Sanscot Technology AB.

SFS 2014:114. Lag om ändring i miljöbalken. Stockholm: Miljödepartementet.

98

http://www.molare.net/disaster/the_disaster_4.html

http://www.molare.net/disaster/the_disaster_4.html

https://www.norconsult.se/referenser/2014/lima-kraftverk/

https://www.norconsult.se/referenser/2014/lima-kraftverk/

http://www.oofem.org/resources/doc/elementlibmanual/html/elementlibmanual.html

http://www.oofem.org/resources/doc/elementlibmanual/html/elementlibmanual.html

http://www.risk-assessment.at/news_en.html

http://www.risk-assessment.at/news_en.html

BIBLIOGRAPHY

Shaffner, P., Scott, G., 2013. The knowledge of precedents and their influence onprofessional development.

Svenska Kraftnät, 2014. Regelverk och riktlinjer. http://www.svk.se/

aktorsportalen/dammsakerhet/regelverk-och-riktlinjer/, retrieved2015-03-03.

The Engineering News Publishing company, 1910. Partial Failure of a Concrete Damat Austin Pa. Engineering News 63, 321–323.

The Engineering News Publishing company, 1911. The Partial Failure of a ConcreteDam at Austin, Pa., on Jan. 23, 1910. Enineering news v.66 (14), 417–424.

Trafikverket, 2011. TK Geo 11 Trafikverkets tekniska krav för geokonstruktioner.Trafikverket.

Vattenkraft.info, 2009. Info om Svensk vattenkraft. http://vattenkraft.info/teori/bilder/ratan.jpg, retrieved 2015-03-03.

Veale, B., Davison, I., 2011. Estimation of gravity dam breach geometry Review ofcurrent practice.

Water and power, 2015. St. francis dam disaster. http://waterandpower.org/

museum/St.%20Francis%20Dam%20Disaster.html, retrieved 2015-04-23.

Westberg, M., 2010. Reliability-based assessment of concrete dam stability. DoctoralThesis TVBK-1039, Division of Structural Engineering, Lund University.

Westberg, M., 2014. Stabilitetsanalys av betongdammar - läget för pågåenge utveck-ling av sannolikhetsbaserad metod. Worshop 25/11 2014 om Eurokoder medtillämpning på betongdammar. Energiforsk(Elforsk), 2014-11-25.

Westberg, M., Hassanzadeh, M., 2007. Stabilitetsanalys av betongdammar. Elforskreport 07:11, Energiforsk(Elforsk), seminar 2006-11-27.

Wilson, D., 1963. Baldwin Hills Reservoir disaster. http://jpg3.lapl.org/

pics21/00060040.jpg, los Angeles Public Library Images, retrieved 2015-05-20.

99

http://www.svk.se/aktorsportalen/dammsakerhet/regelverk-och-riktlinjer/

http://www.svk.se/aktorsportalen/dammsakerhet/regelverk-och-riktlinjer/

http://vattenkraft.info/teori/bilder/ratan.jpg

http://vattenkraft.info/teori/bilder/ratan.jpg

http://waterandpower.org/museum/St.%20Francis%20Dam%20Disaster.html

http://waterandpower.org/museum/St.%20Francis%20Dam%20Disaster.html

http://jpg3.lapl.org/pics21/00060040.jpg

http://jpg3.lapl.org/pics21/00060040.jpg

Appendix A

Compiled failures

101

APPENDIX A. COMPILED FAILURES

Dam name

Co-untry

Type Height lowest found

Year com.

Year failure

Fund mat.

Geology Fail. type

Failure mode

foundation

Fail. mode dam

Failure due to overtopping

Failure Description

P SC S Bayless USA M 17 1909 1910

(1911) Rock Sandstone horizontal

layers with shale and clay between

Ff x Overtopping due to unknown cause

See Section 4.3.1

Camara Brazil M 50 2002 2004 Rock Plane of micaceous silty clay

Ff x Not at highest water level

See Section 4.3.1

Eigiau GB M 10 1911 1925 Clay Hard blue clay containing boulders of granite overlain by a layer of peat

Ff/Fm

x Not at highest water level

See Section 4.3.1

Elwha USA M 51 1912 1912 Soil/ Rock

Fluvioglacial and conglomerate

Ff x Filled See Section 4.3.1

High Falls USA (NY)

M 9 1910 1999 Rock No information Fm ST Overtopping due to unknown cause

Overtopping led to breach of 23 meter long portion of concrete crest cap, left half the spillway. Repairs completed.

Marquette no 3

USA M 10 1924 2003 Rock No information Fa Overtopping due to unknown cause

Overtopping and failure of abutment, due to failure of upstream dam.

Shih-Kang dam

Taiwan

M 22 1977 1999 Rock Top deposition layer: unconsolidated gravel, sands, silts and clay. On Soft bedrock: slate-gray, sandy-shale and silty-sandstones

Ffb EQ No information See Section 4.3.1

St Francis USA M 62 1926 1928 Rock Conglomerate and schist

Ff x x Gradual during first fill

See Section 4.3.1

Torrejon-Tajo

Spain M 62 1967 1965 No info

No information Fa/ Fm

SH Flood during construction

Shear sliding within the dam. Failure cause was traced to organic material present in the aggregate and filling of the dam by a flood during construction before the concrete had fully hardened.

Upriver dam

USA M 11,5 1937 1986 Soil No information Fa x Overtopping due to unknown cause

Washout of the abutment and the power canal embankments due to overtopping. Not a complete failure and reparations of the dam were possible.

Warrens-burg

USA M 8 1909 1976 No info

No information Fa T/C No information Breach of north abutment. Reconstructed in 1998.

Xuriguera Spain M 42 1902 1944 Rock No information Ff x No information Failed by foundation sliding, shear strength and poor design.

Zerbino Italy M 16 1925 1935 Rock Schist and hornfeld Faf x x Overtopping due to unknown cause

See Section 4.3.1

Ashley USA B 18 1908 1909 Soil Fluvioglacial Ff x Just spilling when pipe failed

Piping failure in fine sand with little clay and gravel, 6m deep below cut-off.

Cascade lake dam

USA B 5 1908 1982 Soil Glacial terminal-moraine sediments

Ffa x Overtopping due to unknown cause

The dam was overtopped before tipping over and failing. The cause of failure was the hydrostatic water pressure on the dam and erosion of the abutments. Stored water was released rapidly due to short time of breach development and the width of the breach was large.

Komoro Japan B 16 1927 1928 Rock Tuff Ff x x No suggestions of high water level

Failure due to softening of volcanic ash in foundation. Unclear cause, either piping, sliding or both.

Morris Sheppard

USA B 58 1941 1986 Rock Shale Ff x Releases kept within channel capacity

See Section 4.3.1

Overhol-ser

USA B 17 1920 1923 Rock No information Ffa x Overtopping due to unknown cause

Overtopping leading to scour of abutment.

Stoney creek

USA B 21 1913 1914 Soil No information Ff x Not clear if failed at top level

Piping in foundation followed by settling of dam, cracking and collapse of dam.

102

Appendix B

Results analytical analyses

Table B.1: Safety factors from analytical calculations.

RIDAS Eurocode Eurocode mod.Dam Type Sliding Overturning Sliding Overturning Sliding Overturning

Dam 1 M 0.6666 0.8313 0.9473 0.7689 0.8820 0.6693Dam 2 M 0.4780 1.0043 0.9464 0.7774 0.8887 0.6471Dam 3 B 0.7757 1.9554 1.1092 1.7549 1.0121 1.5157Dam 4 B 0.5648 1.6531 0.9773 1.3381 0.8927 1.1301Dam 5 M 0.3619 1.0329 0.7291 0.7943 0.6871 0.65577Dam 6 B 1.3266 0.9423 2.2461 0.7095 2.1750 0.6133Dam 7 M 0.2875 1.7074 0.4863 1.5023 0.4387 1.1163Dam 8 S 0.8709 1.1749 1.3687 1.0666 1.2402 0.7944Dam 9 S 0.2554 2.0430 0.4904 1.7917 0.4229 1.2301Dam 10 P 0.0407 2.0823 0.1122 1.6909 0.0765 1.3126Dam 11 P 0.5708 1.3136 0.8598 1.1477 0.7974 0.9135Dam 12 S 0.4829 1.7511 0.6297 1.6297 0.6373 1.1884Dam 13 P 0.5911 1.4398 0.9448 1.1677 0.8996 0.9760Dam 14 B 0.8161 1.5763 1.1960 1.3468 1.1024 1.1654Dam 15 P+S 0.6045 1.6667 0.9647 1.4866 0.8971 1.0663Dam 16 P 0.9583 1.4290 1.4689 1.2314 1.3700 0.9785Dam 17 P 0.6262 1.5790 0.9786 1.3772 0.9176 1.0684Dam 18 P 0.7225 1.5493 1.1330 1.3227 1.0657 1.0421

103

APPENDIX B. RESULTS ANALYTICAL ANALYSES

Table B.2: Safety factors from analytical calculations including rock bolts

RIDAS Eurocode Eurocode mod.Dam Type Sliding Overturning Sliding Overturning Overturning

Dam 1 M 0.6666 0.8313 0.9473 0.7689 0.6222Dam 4 B 0.5139 1.7919 0.8002 1.5901 1.3115Dam 6 B 1.0075 1.2350 1.2858 1.1966 0.9719Dam 8 S 0.6873 1.3374 0.8117 1.3743 0.9896Dam 15 P+S 0.5763 1.7377 0.8608 1.6269 1.1520

104

Appendix C

Output values for dams

Table C.1: Dam 1, massive dam, height 5m.

Gravity Water H Ice Uplift V Soil H Soil V Anch x Anch y

RIDAS Loads [MN] 2.63 1.09 2.00 0.90 0.62 0.90 0.31 0.61Lever arm [m] 2.57 1.57 4.38 2.59 1.47 0.96 0.00 3.41

EurocodeLoads [MN] 2.74 1.11 2.00 0.92 0.60 0.86 0.71 1.41Lever arm [m] 2.57 1.57 4.38 2.59 1.47 0.96 0.00 3.41


Gravity Water H Ice Uplift V Soil H Soil V

RIDASLoads [MN] 5.20 2.48 2.00 1.90 1.70 2.50Lever arm [m] 3.62 2.37 6.78 3.63 2.27 1.50

EurocodeLoads [MN] 5.43 2.53 2.00 1.94 1.64 2.39Lever arm [m] 3.62 2.37 6.78 3.63 2.27 1.50

Table C.3: Dam 3, buttress dam, height 40m.

Gravity Water H Ice Uplift V Water V

RIDASLoads [MN] 42.12 55.48 1.60 2.95 34.41Lever arm [m] 17.56 12.53 37.27 34.33 27.16

EurocodeLoads [MN] 49.95 56.55 1.60 3.01 35.08Lever arm [m] 17.56 12.53 37.27 34.33 27.16

105

APPENDIX C. OUTPUT VALUES FOR DAMS


Gravity Water H Ice Uplift V Soil H Soil V Anch x Anch y Water V

RIDASLoads [MN] 5.75 5.10 1.60 0.78 2.31 1.25 0.12 0.54 1.55Lever arm [m] 8.38 3.80 11.07 12.18 3.33 2.38 0.00 12.04 11.68

EurocodeLoads [MN] 6.00 5.20 1.60 0.79 2.11 1.29 0.27 1.23 1.58Lever arm [m] 8.38 3.8 11.07 12.18 3.33 2.38 0.00 12.04 11.68


Gravity Water H Ice Uplift V Soil H Soil V Water V

RIDASLoads [MN] 0.25 0.11 0.10 0.08 0.05 0.08 0.01Lever arm [m] 2.11 0.87 3.80 2.34 1.33 0.67 3.35

EurocodeLoads [MN] 0.26 0.11 0.10 0.08 0.05 0.08 0.01Lever arm [m] 2.11 0.87 3.80 2.34 1.33 0.67 3.35


Gravity Water H Ice Uplift V Anch x Anch y Water V




Gravity Water H Ice Uplift H Uplift V Soil H

RIDAS Loads [MN] 1.65 0.61 0.10 0.06 0.85 0.11Lever arm [m] 7.09 4.57 11.70 0.60 7.19 6.10


Soil H3 Soil V Soil V2 Water V Tailw H Tailw V

RIDAS Loads [MN] 0.15 0.65 0.08 0.19 0.08 0.06Lever arm [m] 2.05 2.73 0.96 1.03 0.00 3.64


106

Table C.8: Dam 8, spillway, height 7m.

Gravity Water H Water H2 Ice Uplift H Uplift V Anch x



Anch y Water V Tailw H Tailw V

RIDAS Loads [MN] 0.08 0.02 0.02 0.02Lever arm [m] 6.35 10.85 0.00 9.88

EurocodeLoads [MN] 0.19 0.02 0.02 0.02Lever arm [m] 6.35 6.85 0.73 0.49


Gravity Water H Uplift V Water V Water V2 Tailw H Tailw V



Tend x Tend x2 Tend y Tend y Tend y2

RIDAS Loads [MN] 0.03 0.09 0.14 0.19 0.07Lever arm [m] 4.67 4.01 7.88 5.88 5.88


Table C.10: Dam 10, pillar, height 13m.

Gravity Water H Water H2 Water H3 Ice Ice 2 Ice 3 Uplift H

RIDASLoads [MN] 6.16 1.59 0.07 0.58 0.56 0.23 0.50 0.59Lever arm [m] 7.12 0.37 8.80 4.60 8.10 8.10 8.10 2.40


Uplift V Soil H Water V Tailw H Tend x Tend x2 Tend y Tend y2

RIDAS Loads [MN] 2.00 0.04 0.08 0.12 0.78 0.15 1.56 1.74Lever arm [m] 7.30 2.60 1085.45 480.03 8.16 5.00 4.35 2.05


107

APPENDIX C. OUTPUT VALUES FOR DAMS


Gravity Water H Water H2 Water H3 Ice Uplift V




Gravity Water H Uplift V Soil H




Gravity Water H Water H2 Ice Uplift V



H.W shutter V.W shutter H weight shutter V weight shutter




Gravity Water H Ice Uplift V Water V



108

Table C.15: Dam 15, pillar and spillway, height 16/2m.

Gravity Water H Water H2 Water H3 Ice



Uplift V Anch x Anch y Water V Tailw H








Gravity Water H Water H2 Ice Uplift V







109

TRITA - BKN. Master Thesis 455, Concrete Structures 2015

ISSN 1103-4297

ISRN KTH/BKN/EX-455-SE

www.kth.se

evaluation of failure modes for concrete...

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