design and construction of a bridge deck using frp as … · a research project was undertaken to...

Università degli Studi di Napoli “Federico II”

Corso di laurea in Ingegneria Civile

Tesi

DDEESSIIGGNN AANNDD CCOONNSSTTRRUUCCTTIIOONN OOFF AA BBRRIIDDGGEE DDEECCKK UUSSIINNGG FFRRPP AASS MMIILLDD AANNDD

PPOOSSTT--TTEENNSSIIOONNEEDD RREEIINNFFOORRCCEEMMEENNTT

Relatore: Ch.mo Prof. Ing. G. Manfredi

Correlatore: Ch.mo Prof. Ing. A. Nanni

Candidato: Raffaello Fico

Matr. 37/1991

Anno Accademico 2003-2004

BRIDGE DECK CONSTRUCTION WITH

INTERNAL FRP REINFORCEMENT

VALIDATION OF FRP COMPOSITE TECHNOLOGY

THROUGH FIELD TESTING

Construction of

Southview Bridge Deck

Prepared for:

University of Naples

April-October, 2004

iii

ABSTRACT

A research project was undertaken to evaluate the use of post-

tensioned FRP for bridge-deck construction.

The type of structure selected for this project is a four-span continuous

concrete slab having GFRP bars for top and bottom mats and CFRP

reinforcement for internal post-tensioning of the bridge deck. This

bridge is located in Rolla, Missouri (Southview Drive on Carter

Creek). One lane of the bridge was already built using a conventional

four-cell steel reinforced concrete box culvert. One lane and sidewalk

needed to be added. This additional lane was constructed using FRP

bars as internal reinforcement.

The combination of prestressed and non-prestressed FRP

reinforcement resulted in an economical solution for a deck system

with low deflection and high shear strength at a minimum deck

thickness.

This study includes the design of the FRP portion of the bridge using

existing codes when appropriate, the validation of the FRP technology

through a pre-construction investigation conducted on two specimens

representing a deck strip, construction and field evaluation through

load testing of the bridge.

The results showed how FRP, in the form of GFRP as passive and

CFRP bars as active internal reinforcement, could be a feasible

solution replacing the steel reinforcement of concrete slab bridges, and

specifically the enhanced shear capacity of the slab due to the CFRP

prestressing.

iv

ACKNOWLEDGMENTS

The first and greatest thanksgiving is referred to my parents, who gave

me the possibility to achieve this wonderful experience; thanks to my

best friends, Stefano, Giuseppe, Annalisa.

I wish to sincerely express thankfulness to Dr. Gaetano Manfredi and

Dr. Andrea Prota, who trusted in my capabilities and gave me the

great opportunity to live a so exciting adventure.

I want to express sincere appreciation to Dr. Antonio Nanni for his

guidance. His advice has been a very important factor that enabled me

to accomplish this research and gives me excitement for the future.

Sincere appreciation is also extended to Dr. Nestore Galati, for his

review of this thesis; I bothered him many times and he never hinted

any sign of nuisance.

This thesis was also a pleasing walking along which I met some

extraordinary people:

Andrea, who spent many hours teaching and helping me setting up

machines and tools in the ERL Lab, besides he was been my favourite

friend in Rolla, living with him whole weeks was like being in a funny

sit-com.

Bill, whose help and support was so selfless and helpful for my

activity, and Kathie, my American mamma.

Jason, whose skills were incredible and so needful for this project; he

helped me in the most difficult moments, behind a rough person there

was a large hearted, honest and lovely friend. He made this project a

special adventure that I will always remember with pride and joy; and

v

Abbie, whose kindness and effectiveness were simply amazing for

me. They are the best guide for my little buddy, Logan.

Thank you to my first Italian fan, Nina, who supported me in the

worst moments of my academical life.

Special appreciation goes to Umberto, who spurred me to accept the

proposal to go to USA for this thesis.

A hearty thank you to the brothers and sisters who upheld me during

my American stay: Fausta, Anna, Mario, Anita, Mariagrazia, Elia,

Anna Valentina, Dario, Marco, Davide, Carmine, Nico, Elena.

A brief but proud recall to the friends that sheared with me the most

intense periods of my “long” academical experience: Giuseppe

Malgeri, Giuseppe Carmignano and Graziano.

…Yet this thesis is dedicated to the person that watched over me more

than everyone: ciao Nonna, grazie …

vi

TABLE OF CONTENTS

ABSTRACT...............................................................................................iii

ACKNOWLEDGMENTS......................................................................... iv

1 INTRODUCTION ............................................................................. 15

1.1 BACKGROUND ................................................................................... 15

1.2 RESEARCH OBJECTIVES................................................................... 16

1.3 DESCRIPTION OF THE PROJECT...................................................... 16

1.4 THESIS OUTLINE................................................................................ 17

2 LITERATURE REVIEW.................................................................. 19

2.1 GENERAL............................................................................................. 19

2.2 INTRODUCTION ................................................................................. 20

2.3 RELATED LITERATURE REVIEW .................................................... 21

2.3.1 Non-Prestressed FRP Reinforced Bridge Decks.......................... 21

2.3.1.1 Bridge Street Bridge, Michigan........................................... 21

2.3.2 Prestressed FRP Reinforced Bridge Decks.................................. 25

2.3.2.1 Evaluation of Frp Prestressed Panels/Slabs for I-225/Parker

Road Project......................................................................................... 25

2.3.3 Research works on Shear Behavior of Prestressed FRP............... 31

2.3.3.1 S.Y. Park and A.E. Naaman (1999)..................................... 32

2.3.3.2 P.A. Whitehead and T.J. Ibell (Jan. 2004) ........................... 34

2.3.3.3 P. A. Whitehead and T. J. Ibell (Feb. 2004)......................... 35

3 EXPERIMENTAL PROGRAM 39

3.1 SPECIMENS LAYOUT......................................................................... 39

3.2 MATERIAL PROPERTIES................................................................... 40

3.3 SPECIMENS DESIGN .......................................................................... 42

3.4 SPECIMENS PREPARATION.............................................................. 45

3.5 TEST SETUP......................................................................................... 50

vii

3.6 TEST RESULTS AND DISCUSSION................................................... 56

3.6.1 Failure Modes............................................................................. 56

3.6.2 Discussion of Test Results .......................................................... 58

3.7 CONCLUSIONS.................................................................................... 61

4 SOUTHVIEW BRIDGE DESIGN................................................... 62

4.1 INTRODUCTION ................................................................................. 62

4.2 ASSUMPTIONS.................................................................................... 63

4.3 STRUCTURAL ANALYSIS ................................................................. 64

4.3.1 Load Combinations .................................................................... 64

4.3.2 Design Truck and Design Lanes ................................................. 64

4.3.3 Deck Analysis ............................................................................ 66

4.3.4 Flexural and Shear Analysis ....................................................... 66

4.3.5 Summary of the Analysis............................................................ 73

4.3.6 Deflections ................................................................................. 75

4.4 DESIGN ................................................................................................ 76

4.4.1 Assumptions............................................................................... 76

4.4.2 Slab Design ................................................................................ 78

4.4.2.1 Flexural Design................................................................... 78

4.4.2.2 Shear Design....................................................................... 80

4.4.2.3 Temperature and Shrinkage Reinforcement......................... 84

4.4.3 Serviceability.............................................................................. 86

4.4.3.1 Crack Width ....................................................................... 86

4.4.3.2 Long-Term Deflections....................................................... 88

4.4.3.3 Slab Creep Rupture and Fatigue .......................................... 89

4.5 LOAD RATING .................................................................................... 89

4.6 SOUTHVIEW BRIDGE DRAWINGS................................................... 92

5 SOUTHVIEW BRIDGE DECK INSTALLATION......................... 95

5.1 INTRODUCTION ................................................................................. 95

5.2 PRE-CONSTRUCTION MEETINGS.................................................... 95

5.3 SUBSTRUCTURE CONSTRUCTION.................................................. 97

viii

5.3.1 Footing and Floor Construction .................................................. 97

5.3.2 Abutments Construction ........................................................... 102

5.3.3 Walls Construction ................................................................... 104

5.4 SLAB CONSTRUCTION .....................................................................107

5.4.1 Bill of Material ......................................................................... 107

5.4.2 Outline of tasks......................................................................... 110

5.4.3 Investigation of the Slab Construction ...................................... 111

5.4.4 Post-tension of the Slab ............................................................ 121

6 CONCLUSIONS.............................................................................. 127

APPENDICES A. MORE DATA RELATED TO SECTION 2……………………………...134

B. LOAD RATING DETAILED CALCULATIONS……………………….155

C. MATERIAL AND CONSTRUCTION SPECIFICATIONS……………..166

ix

LIST OF TABLES

Table 1 - Mechanical Properties of FRP Bars ................................... 42

Table 2 – Comparison Experimental Theoretical Results ................. 58

Table 3 – Parameters for Girder Analysis ......................................... 67

Table 4 - Moment and Shear per Unit Strip (Live and Dead Load). 74

Table 5 - Material Properties .......................................................... 77

Table 6 - Slab Geometrical Properties and Internal FRP

Reinforcement ........................................................................... 80

Table 7 - Steel and GFRP Flexural Design of the Deck (No Prestress

Used) ......................................................................................... 81

Table 8 - Maximum Shear and Moment due to Live Load.............. 91

Table 9 - Rating Factor for the New Slab (Bending Moment)......... 91

Table 10 - Rating Factor for the New Slab (Shear) ......................... 92

Table 11 - Bill of FRP Material .................................................... 107

x

LIST OF FIGURES

Fig. 1 - Views of the Former Bridge ............................................... 17

Fig. 2 - Cross Section of the Bridge after the Expansion................. 17

Fig. 3 - Cross-sectional Details of Structure B (dim. in SI units) .... 22

Fig. 4 - Erection of DT Girders at the Bridge Site........................... 23

Fig. 5 - Installed DT Girders with External CFCC Post-tensioning

Strands....................................................................................... 23

Fig. 6 - Structure B Completed and Ready for Traffic .................... 24

Fig. 7 - I-225/Parker Road Bridge under Construction.................... 25

Fig. 8 - Installation of Precast Panels.............................................. 26

Fig. 9 - Precast Panels Supported on Two Box Girders .................. 26

Fig. 10 - Top Reinforcement........................................................... 27

Fig. 11 - Cast-in-Place Portion of the Deck .................................... 28

Fig. 12 - Panel and Anchorage Details............................................ 29

Fig. 13 - Construction of the Panels................................................ 30

Fig. 14 - Cross-sectional dimensions and helical reinforcement

shapes for................................................................................... 37

Fig. 15 - Reinforcement Layout (all dimensions in mm)................. 40

Fig. 16 - CFRP Tendons Trace (all dimensions in mm) ................... 40

Fig. 17 - Cage Detail ...................................................................... 45

Fig. 18 - Strain Gages Detail .......................................................... 46

Fig. 19 - Reinforcement Cages ....................................................... 46

Fig. 20 - Specimen Curing.............................................................. 47

Fig. 21 - Specimens after Removal of the Formwork...................... 47

Fig. 22 - Prestressing System.......................................................... 48

Fig. 23 - Steel Wedge Anchorage System....................................... 49

Fig. 24 - Drilling of CFRP Bar after Grout Curing ......................... 49

xi

Fig. 25 - Static Scheme for the Flexural Specimen ......................... 50

Fig. 26 - Hyperstatic Scheme.......................................................... 51

Fig. 27 - Flexural Plot..................................................................... 52

Fig. 28 - Static Scheme for the Shear Specimen ............................. 52

Fig. 29 Shear Plot ............................................................................ 53

Fig. 30 - Steel Plate Detail.............................................................. 54

Fig. 31 - Test Setup (all dimensions in mm) ..................................... 55

Fig. 32 - Shear cracks on the Flexure Specimen ............................. 56

Fig. 33 - Concrete Crushing in the Compressive Zone.................... 57

Fig. 34 - CFRP Bars Rupture Due to Kinking................................. 57

Fig. 35 - Failure in the Shear Specimen .......................................... 58

Fig. 36 - Mmeasured Vs. Mtheoretical....................................................... 59

Fig. 37 – Load deflection Diagram for the Longer Span................... 60

Fig. 38 – Comparison Between Experimental and ............................ 61

Fig. 39 - Bridge T0530 ................................................................... 62

Fig. 40 - View of the Existing Bridge ............................................. 63

Fig. 41 - Truck Load on the New Deck........................................... 65

Fig. 42 - Loading Conditions .......................................................... 66

Fig. 43 - Design Truck on the Girder .............................................. 67

Fig. 44 - Design Truck: Possible Loading Conditions..................... 68

Fig. 45 - Design Lane: Possible Loading Conditions ...................... 68

Fig. 46 - Unfactored Bending Moment Diagrams Due to Live Load

.................................................................................................. 69

Fig. 47 - Unfactored Bending Moment Diagrams Due to Dead Load

.................................................................................................. 70

Fig. 48 - Unfactored Shear Diagrams Due to Live Load ................. 71

Fig. 49 - Unfactored Shear Diagrams Due to Dead Load ................ 71

Fig. 50 - Unfactored Moment and Shear Diagrams......................... 73

xii

Fig. 51 - Bending Moment Envelopes (kN-m/m)............................ 74

Fig. 52 - Bridge Deflection Analysis (Live Load)........................... 75

Fig. 53 - Bridge Deflection Analysis (Dead Load).......................... 75

Fig. 54 - Strength Reduction Factor................................................ 79

Fig. 55 - Flexural Capacities of the Bridge ..................................... 79

Fig. 56 - Rationale for Proposed Shear Strength ............................. 83

Fig. 57 - Bridge Internal FRP Reinforcement: Section at Mid-Span85

Fig. 58 - CFRP Prestressing Tendons ............................................. 85

Fig. 59 - Stress and Strain at Service .............................................. 88

Fig. 60 - Wetting of the Basement and Digging of the Weed.......... 98

Fig. 61 - Work Area after the Water Filling.................................... 99

Fig. 62 - Devices to Remove or Avoid the Water ........................... 99

Fig. 63 - Footing Formwork ......................................................... 100

Fig. 64 - Pouring of the Footing.................................................... 100

Fig. 65 - Slump Test and Concrete Cylinders Casting .................. 101

Fig. 66 - Footing after Pouring ..................................................... 101

Fig. 67 - Floor Reinforcement ...................................................... 101

Fig. 68 - Water Overflowing and Casting of the Floor.................. 102

Fig. 69 - Laying of Abutments Reinforcement and Formwork...... 103

Fig. 70 - Use of the Steel Ties Gun............................................... 103

Fig. 71 - Casting of the New Abutments....................................... 103

Fig. 72 - Reinforcement and Formwork of the First Wall ............. 104

Fig. 73 - First Wall ....................................................................... 104

Fig. 74 - GFRP Anchoring Detail ................................................. 105

Fig. 75 - Laying of the GFRP Rebars ........................................... 105

Fig. 76 - Making of the Notch ...................................................... 106

Fig. 77 - Central Wall ................................................................... 106

Fig. 78 - Completion of the Prefatory Part of Southview Project .. 106

xiii

Fig. 79 - Formwork Hangers Details............................................. 111

Fig. 80 - Laying of the Formwork................................................. 112

Fig. 81 - Gluing and Placing of Styrofoam ................................... 112

Fig. 82 - Neoprene Pads and Plastic Chairs Details ...................... 113

Fig. 83 - Laying of Bottom Layer of GFRP Rebars ...................... 113

Fig. 84-The RTI Team with (from down on the left) Mr. Cochran, 114

Fig. 85 - Barrier Design................................................................ 114

Fig. 86 - Top GFRP Layer ............................................................ 115

Fig. 87 - Board Detail ................................................................... 116

Fig. 88 - Wooden Board (the wire indicates the position of the

tendons) ................................................................................... 116

Fig. 89 - Slab Section ................................................................... 117

Fig. 90 - Plastic Ducts Detail ........................................................ 117

Fig. 91 - T Connectors Detail ....................................................... 118

Fig. 92 - Attaching of a Strain Gage ............................................. 118

Fig. 93 - Strain Gages Scheme...................................................... 119

Fig. 94 - Pouring Fragments ......................................................... 119

Fig. 95 - Concrete Cylinders Execution and Slab Leveling.......... 120

Fig. 96 - Poured FRP Bridge Deck ............................................... 120

Fig. 97 - Pulling Machine Before and After the Optimizations ..... 121

Fig. 98 - New Pulling Machine Ready for the Use........................ 122

Fig. 99 - Load Cell and External Chuck Detail ............................. 122

Fig. 100 - Hammer Detail and Use on the Site.............................. 123

Fig. 101 - Orange Box .................................................................. 123

Fig. 102 - Wedges Insertion and Cutting of the Tendon................ 124

Fig. 103 - First Set of Pulled Tendons .......................................... 124

Fig. 104 - Grout Injection and Slots Locking................................ 125

Fig. 105 - Cutting of the Slab Edge .............................................. 126

xiv

Fig. 106 - Southview Bridge-Deck Completed ............................. 126

15

1 INTRODUCTION

1.1 Background

Reinforced and prestressed concrete (PC and RC) structures are facing

a worldwide problem, which is the corrosion of the steel as a result of

aging and aggressive environments. Steel corrosion leads to member

degradation, endangers structural integrity and may even cause

catastrophic failures. Research has been carried out in an effort to find

the solution for this problem. The recent advancements in the field of

material science have resulted in the development of new products

that can be used in many areas of civil engineering where

conventional materials have failed to provide satisfactory service life.

Fiber Reinforced Polymers (FRP) have been proposed for use in lieu

of steel for reinforcement and prestressing tendons in concrete

structures. The promise of FRP materials lies in their high-strength,

lightweight and non-corrosive, non-conducting, and non-magnetic

properties. In addition, FRP manufacturing offers a unique

opportunity for the development of shapes and forms that would be

difficult or impossible with conventional steel materials. They can be

manufactured as reinforcing bars and tendons for RC and PC

structures, sheets and laminates for external strengthening of beams,

slabs and masonry walls, wraps and shells for confinement of

columns, etc.

The interest in the use of FRP composites in prestressed concrete is

mainly based on durability issues. Corrosion of prestressing steel

16

tendons caused serious deterioration of infrastructure. Properties like

high tensile strength and high resistance to corrosion would appear to

make FRP composites good candidates for prestressing tendons.

1.2 Research Objectives

The objectives of the project at the University of Missouri Rolla were

as follows:

1. Evaluate the feasibility, behavior, and effectiveness of the new deck

system, showing how FRP, in the form of GFRP as passive and CFRP

bars as active internal reinforcement, could be an excellent solution

replacing the steel.

2. Provide analytical data in support of the enhanced shear capacity of

the concrete slab due to the CFRP prestressing.

1.3 Description of the project

The City of Rolla in Missouri has made available a bridge (Southview

Drive on Carter Creek) to demonstrate the use of FRP bars and

tendons in new constructions. One lane of the bridge was already

constructed using conventional four-cell steel reinforced concrete

(RC) box culvert. It consists of a steel RC slab about 25 cm (10 in)

thick, as depicted in Fig. 1. The slab deck is continuous over three

intermediate reinforced concrete vertical walls, and the overall length

of the bridge is roughly 12 m (40 ft). The new deck was built on three

conventional RC walls as for the existing structure. The expansion

phase included the removal of the curb from the existing RC deck to

allow extending the overall width of the bridge from 3.9 m (12.8 ft) to

17

11.9 m (39 ft). The curb-to-curb width of the resulting bridge is 9.1 m

(30 ft). The two additions consist of a FRP prestressed/reinforced

concrete deck and a steel RC deck as shown in Fig. 2. The

construction of the bridge started on July 2004 and finished on

October 2004.

Fig. 1 - Views of the Former Bridge

WEARING SURFACE5 cm ASPHALT

EXISTING DECKCURB FROM

REMOVE EXISTING

BARRIER WALLCONCRETE

REINFORCED

SIDEWALK (1.5m)NEW CONCRETE

CONCRETE DECKREINFORCED

STEEL

FRP REINFORCED CONCRETE DECK EXISTING REINFORCED CONCRETE

4.3 m1.5 m

12 m2 m3.9 m6 m

Fig. 2 - Cross Section of the Bridge After the Expansion

1.4 Thesis outline

This thesis consists of six sections:

• Section one gives a brief introduction of the background of FRP

applications in civil engineering and introduces the objectives

of the research.

• Section two contains informations on existing prestressing and

non-prestressing FRP bridge decks, so that they can be

compared with the new deck system that is the subject of this

18

thesis. It also gives a summary of the main research works on

shear behavior of prestressed FRP.

• Section three presents the pre-construction investigations

conducted on two specimens representing a deck strip 457 mm

(18 in) wide and 7 m (23 ft) long, fabricated and tested. The

specimens were constructed by the contractor peculiar for the

project allowing for his familiarization with the use of non-

conventional materials. The testing of the specimens as

continuous slabs over three supports allowed the validation of

the design calculations in terms of flexure and shear capacities.

• Section four focuses on Southview Brigde design, providing the

structural analysis of the new FRP concrete bridge deck based

on the AASHTO HS20-44 design truck and providing

calculations for its design using a combination of non-

traditional corrosion-resistant composites materials.

• Section five details the installation of Southview bridge deck,

focusing on the post-tensioning of the slab, the most

considerable and crucial part of the project.

• Section six gives a brief summary of the steps developed in the

previous sections and deepens the results of the research work

and installation of the bridge-deck.

19

2 LITERATURE REVIEW

2.1 General

There are only about fifty bridges in the world with Fiber Reinforced

Polymer superstructure as reported by the U.S. Department of

Transportation Federal Highway Administration1 as of July 2002.

The use of FRP bridge decks has been a direct result of technology

transfer initiatives taken by the defense industry in late 1980’s and

early 1990’s. In fact, many of the manufacturers of FRP bridge decks

were directly or indirectly related to the aerospace composites

industry. Furthermore, in recent years many state transportation

departments in partnership with the Federal Highway Administration

(FHWA) are introducing these new materials to bridge structures with

the intention of gaining design and construction experience and long-

term performance data on FRP, thus FRP deck systems are emerging

as a viable alternative to conventional systems, namely reinforced-

concrete slabs. Moreover the research on prestressing with FRP

tendons is getting attention mainly due to the fact that nearly half of

the nation’s bridges are reported to be in serious disrepair or

functionally obsolete.

The use of such systems to replace existing, deteriorated bridge deck

systems offers both economic benefits and improved performance.

The economic advantages are possible for a number of reasons: since

such composite systems are lighter, considerable savings are realized

20

by reduced transportation costs (several deck systems can be

transported on one truck); erection costs will be less as relatively light

cranes can be used to install the decks; and construction time is

reduced, which eliminates long traffic delays. Due to the high

resistance of FRP deck systems to environmental effects and corrosion

attack, the long term performance is also expected to be improved

significantly, leading to lower maintenance and longer service life. In

addition to economic advantages, FRP deck systems offer structural

advantages as well. For example, higher live loads can be resisted by

supporting steel stringers, as the dead load applied by the FRP deck

system is about one fifth of a conventional reinforced-concrete deck.

2.2 Introduction

The proposed literature review refers to different topics related to the

concrete bridge decks using FRP reinforcement:

• Non-prestressed FRP reinforced bridge decks.

• Prestressed FRP reinforced bridge decks.

• Research works on shear behavior of prestressed FRP.

In this section, some of the main existing internal FRP bridge decks

have been presented, so that they can be compared with the new deck

system that is the subject of this thesis.

Other existing samples are in Appendix A.

21

2.3 Related Literature Review

2.3.1 Non-Prestressed FRP Reinforced Bridge Decks

2.3.1.1 Bridge Street Bridge, Michigan

The Bridge Street Bridge in Southfield, Michigan, is the first vehicular

concrete bridge ever built in the United States that uses CFRP material

as the principal structural reinforcement. The project consists of two

parallel bridges - Structures A and B – over the Rouge River in the

City of Southfield. Both structures use three skewed spans, each over

62 m (204 ft) long, to carry vehicular traffic. Structure A consists of a

new substructure as well as a new superstructure, and incorporates

five equally spaced conventional AASHTO Type III girders in each of

its three spans. Its cast-in-place concrete deck slab is placed

continuously across the three spans. Structure B consists of four

special precast, prestressed double-tee (DT) girders in each of the

three spans configured as simply supported spans. Each DT girder is

structurally reinforced polymer (CFRP) LeadlineTM tendons and post-

tensioned carbon fiber composite cable (CFCC)TM strands in both

longitudinal and transverse directions. The non-prestressed

reinforcement in the girders and deck structure consists of CFCC

strands manufactured in bent configurations, straight CFCC

reinforcing bars, CFRP NEFMACTM grid reinforcement, and stainless

steel reinforcing bars for stirrups.

The bridge cross section (see Fig. 3) consists of four precast DT

sections and a minimum 75 mm (3 in.) thick non-continuous deck

slab.

22

Fig. 3 - Cross-sectional Details of Structure B (dim. in SI units)

The composite topping is reinforced with NEFMAC grids. The

composite section is also reinforced with four externally draped 40

mm (1.57 in.) diameter unbonded CFCC post-tensioning strands.

The unique construction method was based on a process developed

and tested previously at LTU. The transportation of the 12 girders

from the precast plant in Windsor, Ontario, to the bridge site in

Southfield, Michigan, required a special barging arrangement. The

girders were erected using two large capacity cranes at opposite ends

of the bridge.

Fig. 4 shows the erection of a DT girder for the south span, and Fig. 5

shows the installed girders with external post-tensioning strands in

place.

23

Fig. 4 - Erection of DT Girders at the Bridge Site

Fig. 5 - Installed DT Girders with External CFCC Post-tensioning Strands

The Bridge Street Bridge Deployment Project has served as an

extraordinarily successful example of technology transfer from

research and development to serviceable structure.

The bridge exhibits innovation not only in the material itself, but also

in the variety of prestressing methods implemented – pretensioning

and post-tensioning, internal and external.

This project recently won PCI’s Harry H. Edwards Industry

Advancement Award in the recent PCI Design Awards Program (see

Fig. 6).

24

Fig. 6 - Structure B Completed and Ready for Traffic

25

2.3.2 Prestressed FRP Reinforced Bridge Decks

2.3.2.1 Evaluation of Frp Prestressed Panels/Slabs for I-225/Parker Road Project

Under the Innovative Bridge Research and Construction (IBRC)

program of the Federal Highway Administration (FHWA), in 2001 the

Colorado Department of Transportation (CDOT) has introduced fiber

reinforced polymeric (FRP) reinforcement in a bridge project at I-

225/Parker Road. Precast prestressed concrete panels were used as

stay-in-place forms for a bridge deck. The bridge during construction

is shown in Fig. 7.

Fig. 7 - I-225/Parker Road Bridge under Construction

Some of these panels were prestressed with CFRP tendons and the rest

with regular seven-wire steel strands. The primary objective of this

project was to demonstrate the feasibility of using CFRP tendons in

place of seven wire steel strand for prestressed concrete panels.

26

It was the first time CFRP bars were used in such fashion. The precast

panels were supported on two cast-in-place, post-tensioned concrete

box girders (Fig. 8 and Fig. 9). A topping slab was added to the panels

to form a composite bridge deck to carry the traffic.

Fig. 8 - Installation of Precast Panels

Fig. 9 - Precast Panels Supported on Two Box Girders

Furthermore, the applicability of current AASHTO provisions to

CFRP prestressed panels was investigated. In addition to the

experimental investigation, a new rational design methodology for

27

bridge decks was proposed and investigated (Borlin 2001) in this

project. Most highway bridges in the United States have slab-on-

girder decks, in which steel or precast concrete girders support a

reinforced concrete slab. The two components are tied together with

shear connectors to allow for composite action. The main

reinforcement in the deck slabs is placed perpendicular to the direction

of traffic. For these decks, both the AASHTO Standard Specifications

for Highway Bridges (AASHTO 1996) and the conventional method

in the AASHTO LRFD Bridge Design Specifications (AASHTO

1998), resulted in two dense reinforcement mats, one in the top of the

slab and one in the bottom (Fig. 10 and Fig. 11).

Fig. 10 - Top Reinforcement

28

Fig. 11 - Cast-in-Place Portion of the Deck

Pullout tests conducted in this study showed that Leadline CFRP

prestressing tendons and C-BAR glass fiber reinforced polymeric

(GFRP) reinforcing bars had higher bond strengths than seven-wire

steel strand and regular steel reinforcing bars, respectively.

Load tests were first performed on two panels, one prestressed and

reinforced with FRP and the other prestressed and reinforced with

steel. Both panels were designed to barely satisfy the AASHTO

specifications. An additional panel that was prestressed and reinforced

with FRP and brought from the I-225/Parker Road project was tested

as well. This panel was conservatively designed with a significant

reserve capacity compared to the first two. Load tests were also

performed on steel and FRP prestressed panels that had a 125 mm (5

in.) composite topping slab (see Fig. 12 and Fig. 13). All test results

showed the feasibility of using CFRP tendons for prestressing and

GFRP bars for temperature and distribution reinforcement in precast

bridge panel construction. The GFRP reinforcement was selected

29

according to the recommendation for temperature reinforcement in the

ACI 440H draft report.

Load distribution data were taken during the composite slab tests to

validate the adequacy of the Equivalent Width Strip method used in

AASHTO LRFD Specifications. The method was found to be

conservative for both steel and FRP reinforced composite slabs.

However, results showed that the steel reinforced slab better

distributed the loading in the transverse direction than the FRP

reinforced slab. This suggests that the recommendation for

temperature reinforcement in the ACI 440H draft report may not be

adequate for distribution reinforcement.

Fig. 12 - Panel and Anchorage Details

30

Fig. 13 - Construction of the Panels

31

2.3.3 Research works on Shear Behavior of Prestressed FRP

The majority of research on concrete structures using FRP

reinforcement has been on members that are not critical in shear tests.

At present, the shear behaviour of prestressed concrete members using

FRP reinforcement is not well understood. Unlike flexural behavior,

shear behaviour is quite complex itself, even in ordinary reinforced or

prestressed concrete members.

Structures are usually conservatively designed, and rely on plasticity

theory for safety. This ensures that, if a set of internal forces exists

which is in equilibrium with the applied load, and since it is known

that steel is ductile, the lower bound (or ‘Safe Load’) theorem can be

used to assert that the structure is safe.

When advanced composites are used, however, the theoretical

justification is much less sound; many of the basic assumptions no

longer hold. Composites are generally less stiff than steel so, when the

concrete cracks, a composite is carrying less force than steel would be.

Cracks will, thus, be wider, so there will be less concrete–concrete

interaction across the crack; there will thus be less ‘aggregate

interlock’. Composites also delaminate when placed across shear

cracks, so ‘dowel action’ will be lower and there are problems caused

by the bends in the bars.

Finally, and most importantly, although composites have high strain

capacities, they do not behave plasticly, so the Safe Load theorem

cannot be used to hide the lack of knowledge about the deflections.

Taken together, these results mean that care must be taken about

producing design guidelines for shear in compositely reinforced

structures.

32

There is clearly much work to be done in this field—a model is

required which satisfies all three of the basic principles of structural

mechanics—equilibrium, compatibility and the material stress–strain

behaviour [C.J. Burgoyne, August 2001].

Here a summary of the main research works on this topic is given.

2.3.3.1 S.Y. Park and A.E. Naaman (1999)

The authors conducted an experimental program on two series of tests

(two sets of beams). The first series comprised nine prestressed

concrete beams fabricated without stirrups. Five beams were

prestressed using CFRP tendons and, for comparison, four beams

were prestressed using conventional steel tendons.

The main objective of this first series of tests was to experimentally

confirm the shear-tendon rupture failure mode in prestressed concrete

beams with FRP tendons and to compare it with other failure modes in

prestressed concrete beams with steel tendons.

The second series of the experimental program comprised seven FRP

prestressed concrete beams and one non-prestressed concrete beam

shear reinforced with steel stirrups (seven beams) or steel fibers (one

beam). The test parameters were the pretensioning ratio, the shear

span-to-depth ratio, shear reinforcement ratio, the use of steel fibers,

the compressive strength of concrete, and the type of reinforcement.

The main goal of the second series was to evaluate the parameters

affecting the shear strength and ductility of concrete beams

prestressed with FRP tendons.

On the basis of this experimental investigation, the following

conclusions were drawn:

33

1. The shear-tendon rupture failure is a unique mode of failure,

which, unless properly designed for, is likely to occur in

concrete beams prestressed with FRP tendons. This premature

failure is due to tendon rupture by dowel shear at the shear-

cracking plane. It is attributed to the poor resistance of FRP

tendons in the transverse direction and their brittle behavior.

2. The ultimate shear resisting capacity of beams prestressed with

FRP tendons was about 15 percent less than that of beams

prestressed with steel tendons, regardless of their shear failure

mode.

3. The shear-tendon rupture failure occurred at the flexural-shear-

cracking plane in beams with FRP tendons, even when the

effective prestress ratio was low (about 40 %) and the required

amount of steel stirrups was provided according to the ACI

Code.

4. Adding steel fibers is a possible way to improve the shear

resistance of concrete beams prestressed with FRP tendons by

avoiding or delaying shear-tendon rupture failure.

5. Differences in the properties of FRP and steel tendons appear to

have no significant effect on the initial portion of load-

deflection response of prestressed concrete beams subjected to a

center point loading with a shear span-to-depth ratio of 2.5.

6. The ultimate shear displacement and crack width of prestressed

beams that failed by shear-tendon rupture were about one-third

and one-half, respectively, of those of similar beams with steel

tendons.

7. The following observations were made for beams prestressed

with FRP tendons:

34

• Increasing the shear span-to-dept ratio from 1.5 to 3.5 led to a

decrease in shear resistance but an increase in shear ductility

(displacement).

• Adding stirrups in sufficient quantity changes the failure mode

from shear-tension to shear-tendon rupture in beams with a low

effective prestress ratio of about 40 percent.

• Increasing the compressive strength of concrete slightly

increases the shear strength and considerably increases the

corresponding deflection.

2.3.3.2 P.A. Whitehead and T.J. Ibell (Jan. 2004)

The authors developed a model incorporating force equilibrium and

compatibility of strains so that the elastic properties of FRP could be

included rationally.

Fifteen specimens from the experimental program and four case-study

specimens (all of which failed in shear) were used to assess the

accuracy of predictions from the most prevalent codes and design

guides currently in use in the UK and the US, namely BS8110 (1997),

BD44 (1995), EC2 (1992) and ACI-440.1R-03 (2003). ACI-440.1R-03

is specifically intended for FRP-reinforced concrete. However,

predictions obtained using modified versions of the other three

standard codes were also provided. These modifications were termed

the Strain Approach and the Sheffield Approach (Guadagnini et al.,

2001).

The following conclusions were drawn:

1. The analytical shear predictions developed by the authors

predicted the experimental results with a better accuracy when

compared to the existing building codes.

35

2. Using either the Strain or Sheffield modification suggestions in

tandem with BS8110, BD44 and EC2 results in reasonable predictions

for all the reinforced (AFRP and GFRP) specimen capacities. ACI-

440.1R-03 provides more conservative predictions for the FRP-

reinforced specimens.

3. The presence of prestress was found to be significant in

increasing shear capacity of such specimens, due to the significant

crack-closing influence of the prestress. The unfactored code

evaluations are more conservative for the prestressed specimens,

implying that the presence of prestress aids the FRP-reinforced

concrete beams to a greater extent than it does steel-reinforced

concrete.

2.3.3.3 P. A. Whitehead and T. J. Ibell (Feb. 2004)

A novel FRP shear reinforcement strategy, in which both the concrete

and FRP are employed to maximum advantage was conceived by the

authors.

They presented the findings of research conducted into the shear

behaviour of FRP-reinforced and -prestressed concrete beams

containing continuous FRP helical transverse reinforcement. Twelve

tests were conducted on ordinarily-reinforced beams and fifteen on

FRP-prestressed concrete beams.

Tests on FRP-reinforced concrete beams were indeed conducted

initially, thereby allowing rapid assessment of the influence of various

shear reinforcement strategies which could later be investigated under

prestressed conditions. Accordingly, the more effective forms of shear

reinforcement were taken forward and re-examined under prestressed

conditions while the non-responsive forms were discarded. Further, in

36

order to make direct comparison between FRP-reinforced and -

prestressed beams, the effective depth was kept the same for both sets

of tests.

Ibell and Burgoyne suggested that geometry and bond (including the

locality of bond) are of paramount importance in the performance of

FRP shear reinforcement. Therefore, to examine these aspects, the

following FRP shear reinforcement strategies were employed within

the FRP-reinforced and -prestressed rectangular concrete beams:

1. Circular helix, fully-bonded or entirely unbonded, placed along

the top of the beam within the flexural compression zone. See

Fig. 14-a.

2. Circular helix, fully-bonded or entirely unbonded, angled in the

shear zones, following the lines of principal compression. See

Fig. 14-b.

3. Continuous rectangular draped helix, fully-bonded,

intermittently-bonded or entirely unbonded, placed over the full

depth of the section. See Fig. 14-c.

4. Two interlocking rectangular draped helices, entirely unbonded,

placed over the full depth of the section. See Fig. 14-d.

37

Fig. 14 - Cross-sectional dimensions and helical reinforcement shapes for

(a) Specimens containing a circular helix placed along the top of the beam,

(b) Specimens containing a circular helix angled in the shear zone,

(c) Specimens containing a continuous draped rectangular helix, and

(d) A specimen containing two interlocking rectangular helices

The following conclusions were made on shear behavior:

• The presence of prestress was significant in increasing shear

capacity of such specimens.

• When used to resist shear, fully-unbonded circular and

rectangular helices had to be spaced at a closer pitch in

comparison with fully-bonded or intermittently-bonded

rectangular helices to provide a similar increase in failure

capacity. The fully-unbonded rectangular helices appeared to

38

have been about 50% as effective as fully- or intermittently-

bonded rectangular helices in resisting shear.

• Of the arrangements considered, the best technically seemed to

involve the use of a fully-bonded circular helix and a fully-

bonded rectangular helix in the constant-moment region,

coupled with an intermittently-bonded rectangular helix in the

shear zones. This configuration led to considerable

deformability and ductility and produced high shear capacity

and genuine plastic-based ductility during shear collapse.

• FRP reinforcement need not simply be treated as a direct

substitute for steel reinforcement, but that rather its inherent

advantages should be exploited using rational design

approaches.

39

3 EXPERIMENTAL PROGRAM

This section presents the pre-construction investigations conducted on

two specimens representing a deck strip 457 mm (18 in.) wide, 254

mm (10 in.) deep and 7 m (23 ft) long, fabricated and tested. The

specimens were constructed by the contractor peculiar for the project

allowing for his familiarization with the use of non-conventional

materials. The testing of the specimens as continuous slabs over three

supports allowed the validation of the design calculations in terms of

flexure and shear capacities.

3.1 Specimens Layout

Two specimens having the same geometry and amount of

reinforcement were built and tested, one to investigate the flexural

behaviour, the other one the shear behaviour. The specimens were

reinforced using 3 f19 (6/8 in) GFPR bars as top and bottom mat and

2 f9 (3/8 in) CFRP bars as prestressed tendons. The position of the

prestressed tendons was varied along the slab in order to match the

moment demand. In addition, in order to reproduce the actual field

conditions also f13 (4/8 in) GFRP bars spaced 305 mm (12 in) on

center were placed in the transversal direction as temperature and

shrinkage reinforcement. Fig. 15 shows a detailed layout of the

reinforcement, while Fig. 16 shows the position of the CFRP tendons.

40

φ19 GFRP Bars

Temperature ans Shrinkage Reinforcement: φ13 GFRP Bars410 mm Long, 305 mm on Center

φ9 CFRP BarsPRESTRESSING TENDONS

PLASTIC TIES

CHAIRS

38

φ19 GFRP BarsSpaced 152 mm on Center

457

178

38

254

Fig. 15 - Reinforcement Layout (all dimensions in mm)

13706101370127 178178127

178

305 76268668612712717876

76

7010

1067

914 3685 1829 609

152

Supports

Tendon Trace

Fig. 16 - CFRP Tendons Trace (all dimensions in mm)

3.2 Material Properties

Tests were performed to characterize the mechanical properties of the

materials used in this investigation.

The designed concrete compressive strength was equal to 41.4 MPa

(6000 psi). Water to cement ratio for the concrete mixture was 0.45.

The components in the concrete mixture were proportioned as follow

by weight: 19% portland cement, 40% crushed limestone, 33% sand,

and 8% water.

41

The actual compressive strength of the concrete was checked on three

100 x 200 mm (4 x 8 in) cylinders per slab. The cylinders were tested

in compliance with ASTM C39/C39M. The average compressive

strength at the time of the test was found equal to 48.6 MPa (7040 psi)

and 45.4 MPa (6585 psi) for the flexural and shear specimens,

respectively.

The compressive strength of the high performance cementitious grout

was determined on three cylinders 100 x 200 mm (4 x 8 in) and it was

found to be 21.7 MPa (3150 psi) after one day, 36.5 MPa (5300 psi)

after three days, 49.3 MPa (7150 psi) after seven days, and 58.9 MPa

(8550 psi) after 28 days. In addition, splitting tensile tests in

compliance with ASTM C496 were performed on the same type of

cylinders (three repetitions). The splitting tensile strength was found

to be 1.5 MPa (218 psi) after one day, 2.8 MPa (406 psi) after three

days, 3.58 MPa (520 psi) after 7 days, and 5.59 MPa (810 psi) after 28

days.

Tensile tests were performed on FRP bars to determine their

mechanical properties which are related to fiber content. The average

tensile strength, ultimate strain and modulus of elasticity obtained

from the testing of the specimens (ASTM D3039) are presented in

Table 1.

42

Table 1 - Mechanical Properties of FRP Bars

Bar Type

Nominal Diameter of

the Bar mm (in)

Average Max. Strain

%

Average Max. Stress

MPa (ksi)

Average Elastic Modulus GPa (ksi)

GFRP Bar 12.7 (0.5) 1.68 689.5 (100.0) 40.80 (5920)

GFRP Bar 19.1 (0.75) 1.52 620.5 (90.0) 40.80 (5920)

CFRP Bar 9.5 (0.375) 1.67 2124.2 (308.1) 142.7 (20702)

3.3 Specimens Design

The theoretical moment and shear capacities have been computed

according to ACI 440.1R-03 provisions. As an alternative method to

compute the shear capacity of the specimens the equation developed

by Tureyen A. K. and Frosh R. J. and now under consideration for

adoption by ACI Committee 440, was used.

According to the ACI 440.1R-03 the nominal flexural capacity of an

FRP reinforced concrete member can be computed as follows:

1 1( ) ( )2 2n f fu fp fup pc cM A f d A f dβ β

= − + − (3.1)

where:

fA = area of FRP reinforcement

fuf = design tensile strength of FRP, considering reductions for

service environment

d = distance from extreme compression fiber to centroid of tension

reinforcement

1β = factor depending on the concrete strength, 'cf , equal to 0.75 for

'cf = 41.4 MPa (6000 psi)

43

c = distance from extreme compression fiber to the neutral axis

The second term symbols represent the same factors but referred to

the prestressed reinforcement.

The theoretical moment capacity is 97.9 kN-m (72.2 kip-ft), while

without the post-tension reinforcement this value becomes 76.0 kN-m

(55 kip-ft).

According to the ACI 440.1R-03 the concrete shear capacity Vc,f of

flexural members using FRP as main reinforcement can be evaluated

as shown below. The proposed equation accounts for the axial

stiffness of the FRP reinforcement (AfEf) as compared to that of the

steel reinforcement (AsEs).

,

f fc f c

s s

A EV V

A E=

(3.2)

Vc being the nominal shear strength provided by concrete with steel

flexural reinforcement, for members with effective prestress force not

less than 40 percent of the tensile strength of flexural reinforcement,

according to ACI 318R-99:

'(0.6 700 )u pc c p

u

V dV f bd

M= + (3.3)

where: '

cf = specified compressive strength of concrete

uV = factored shear force at section

pd = distance from extreme compression fiber to centroid of

prestressed reinforcement (but needs not to be less than 0.80h for

circular sections and prestressed members)

uM = factored moment at section

44

b = web width

When assuming the CE (environmental reduction factor) equal to 1,

the theoretical shear capacity will be 69.8 kN (15.7 kip).

According to Tureyen A. K. and Frosh R. J. the shear capacity can be

derived from the following equation:

= `, 5c f cV f bc (3.4)

where c is the position of the neutral axis at the service conditions, b

is the width of the specimens and `cf the compressive strength of the

concrete. The shear capacity computed according to this approach is

97.4 kN (21.9 kip).

Indeed using the neutral axis at the service conditions may not be

correct when the member is approaching its flexural capacity; hence,

assuming c corresponding to the ultimate conditions of the section,

the shear capacity will be 54.7 kN (12.3 kip).

The prestressing was primarily used to increase the shear capacity of

the slabs rather than the flexural one. In fact, the shear capacity of the

slabs without any prestressing load would have been 30.7 kN (6.9 kip)

and 33.8 kN (7.6 kip) using the ACI and the Tureyen A. K. and Frosh

R. J., respectively, versus 69.8 kN and 97.4 kN with post-tensioned

reinforcement, respectively. Hence, according to ACI, the post-tension

led to an increase in flexure of ~29%, while the shear is more than

doubled.

45

3.4 Specimens Preparation

The specimens were built in the way to reproduce the same

characteristics of the bridge deck that would be constructed in the

field.

The CFRP tendons were housed inside a plastic duct, in order to allow

the post-tensioning operations. Plastic T joints were used to connect

the duct housing the tendons with the duct going out of the specimen

to inject the grout.

Plastic chairs and ties were used to lay the bars and the tendons in

order to have a completely “steel free” structure, in compliance with

the requirements of the Southview bridge project. Fig. 17 shows the

cage details of both specimens.

Fig. 17 - Cage Detail

A total of 21 Electrical Strain Gages (ESG) were used to monitor the

strain at the most critical sections. They were placed on each GFRP

bar (see Fig. 18) and on the compressive face of the slab. Furthermore

Fig. 19 shows the two specimens ready for pouring.

46

Fig. 18 - Strain Gages Detail

Fig. 19 - Reinforcement Cages

The prestressing of the tendons was executed 28 days after the

pouring of the concrete. Fig. 20 and Fig. 21 show the specimens while

curing and after the removal of the formwork.

47

Fig. 20 - Specimen Curing

Fig. 21 - Specimens after Removal of the Formwork

The CFRP bars were prestressed by applying a force of 98 kN (22 kip)

using hydraulic jacks at both ends. This level of prestressing

corresponds to 65% of the ultimate capacity of the CFRP bars. Such

pre-stressing level was chosen in order to respect, after the initial

strain losses (supposed to be 35% of the initial strain of prestress

reinforcement), the creep rupture limits dictated for GFRP bars

according to ACI 440.1R-03 (0.20 times the ultimate guaranteed

48

tensile strength for prestress CFRP reinforcement), as underlined in

the next section.

Initially the prestressing load was applied only to one end of the slab

causing the breaking of the FRP tendons due to the high eccentricity

of the active reinforcement and to the friction between ductwork and

tendons. This problem was solved by applying the prestressing load in

steps of 31 kN (7 kip) from both ends by mean of two hydraulic jacks

(See Fig. 22). The prestressing load was monitored using two load

cells, one for each end of the slab, while the strain was measured

using two electrical strain gages attached on the bar. This solution was

suitable because the increased losses after the release of the tendons

induced by the new pre-stressing system (30%) were less than the

ones assumed for design (35%).

Anchorage System

CFRP Tendon

Steel Frame

Steel PlateSteel Plates Hydraulic Jack

Anchorage SystemsLoad Cell

End of the Slab

a) System Detail b) Field Assembly

Fig. 22 - Prestressing System

The steel wedge anchorage system used to anchor the CFRP bar and

to react against the hydraulic jack was a resin-free three part system

developed at the University of Waterloo, Canada. It included an outer

steel cylinder, a four-piece wedge and an inner sleeve (see Fig. 23).

The inner sleeve was made out of copper/steel and it was deformable.

The four-piece wedge was placed evenly around the inner sleeve and

49

inserted into the outer steel cylinder. The anchorage system was later

secured by tapping the inner sleeve and four-piece wedge into the

outer steel cylinder with a hammer.

Fig. 23 - Steel Wedge Anchorage System

The prestressing of the CFRP bars was followed by their grouting

using a high performance cementitious grout. The cementitious grout

was allowed to cure for seven days after which the anchoring of the

tendons was removed by drilling the CFRP bar inside the barrel (see

Fig. 24). The strain in the slab was monitored during and for 48 hours

after the cutting of the anchoring system: during this time no loss of

compressive strain in the specimens was recorded.

Fig. 24 - Drilling of CFRP Bar after Grout Curing

Outer steel cylinder

Four-piece Wedge

Inner sleeve

CFRP BarOuter steel cylinder

Four-piece Wedge

Inner sleeve

CFRP Bar

50

3.5 Test Setup

Each slab was tested as a continuous member on three supports,

comprising of a 3.6 m (12 ft) and 1.8 m (6 ft) spans. The positions of

the two loading points were chosen such to force flexural and shear

failure for the flexural and shear specimens, respectively.

They were placed at the mid-span for the flexural specimen:

Fig. 25 - Static Scheme for the Flexural Specimen

Given: 1 3.6 (12 )L m ft= Length of the first span

2 1.8 (6 )L m ft= Length of the second span

1 165 (37 )P kN kip= First live load

2 100 (23 )P kN kip= Second live load (higher than the theoretic value)

Solving the hyperstatic scheme (see Fig. 26) it’s possible to derive the

flexural moment acting on the central support (assigned as hyperstatic

unknown):

51

ϕϕ

Fig. 26 - Hyperstatic Scheme

φBL = φB

R

2 21 1 2 2

1 2

( * * )3 *16

P L P LML L

+= −

+ (3.5)

Hence the reactions of each support can be derived: 1

11

13.22P MR kip

L= − = (3.6)

1 22

1 2

462 2P PM MR kip

L L= + + + = (3.7)

23

2

0.82P MR kip

L= − = (3.8)

Finally the equation of flexural moment and its plot (see Fig. 27)

depending on z can be derived:

1 1( ) *M z R z= (3.9) when 10;2L

z ∈

2 1 1( ) ( ) *( )M z M z P z x= − − (3.10) when 1

1;2L

z L ∈

3 2 2 1( ) ( ) *( )M z M z R z L= + − (3.11) when 2

1 1;2Lz L L ∈ +

4 3 2( ) ( ) *[ ( )]tM z M z P z L y= − − − (3.12) when 2

1 ;2 tL

z L L ∈ +

52

0 2 4 6 8 10 12 14 16 18100

50

0

50

100

Axes of Structure [ft]

Mom

ent [

kip*

ft]

64.125

78.938−

M z( )−

180 z

Fig. 27 - Flexural Plot

max ( ) 86.9 * ( 64.1 * )M z kN m kip ft− = − −

max ( ) 107 * (78.9 * )M z kN m kip ft+ =

Regarding the shear test, the loading points were placed 0.9 m (3 ft)

away from the central support and the distance was determined by

performing the following calculations (See Fig. 28):

Fig. 28 - Static Scheme for the Shear Specimen

According to the same procedure used to solve the flexural scheme,

and given:

53

1 222 (50 )P kN kip= First live load

2 222 (50 )P kN kip= Second live load

the shear equations depending on z and the corresponding plot (see

Fig. 29) where derived: 1 1V R= (3.13) when 0;2.7z m∈

2 1 1V V P= − (3.14) when 12.7 ;z m L∈

3 2 2V V R= + (3.15) when 1 1; 0.9z L L m∈ +

4 3 2V V P= − (3.16) when 1 0.9 ; tz L m L∈ +

0 2 4 6 8 10 12 14 16 1840

20

0

20

40

60

Axes of Structure [ft]

Shea

r [ki

p]

44.531

39.063−

V z( )−

180 z

Fig. 29 Shear Plot

max ( ) 198 ( 44.5 )V z kN kip− = − −

max ( ) 173.5 (39 )V z kN kip+ =

The load was applied in cycles of loading and unloading. An initial

cycle for a low load was performed on each specimen to verify that

both the mechanical and electronic equipment was working properly.

54

Loads were applied to 102 x 457 x 25 mm (4 x 18 x 1 in) steel plates

resisting on the slab in order to prevent that the structure could touch

the edge of the support in the case of large deflections (see Fig. 30).

The flexural specimen (See Fig. 31 –a) was instrumented using three

load cells: two of them were placed under the loading points while the

third one was placed under one of the supports in order to determine

the end reaction and therefore the actual distribution of moments in

the slab.

Fig. 30 - Steel Plate Detail

The loads were applied by means of 30 ton (66 kip) hydraulic jacks

reacting against a steel frame. The loading rate was the same for the

two spans until reaching 85 kN (19 kip). After that, the load in the

shorter span was kept constant while the one in the longer one was

increased up to failure. This solution was adopted in order to avoid

shear failure at the central support. Linear Variable Displacement

Transducers (LVDTs) were positioned at the loading points (two for

each loading point) and at the supports in order to record maximum

displacements and support settlements. The strain gages were placed

on each GFRP bar at the location of the loading points and of the

55

central support. In addition, at the same locations, an additional strain

gage was attached on the compressive face of the slab in order to have

an additional backup point while determining the experimental

moment-curvature response of the slab.

For the shear specimen, the number of load cells was reduced to two.

The loads were applied to 102 x 457 x 25 mm (4 x 18 x 1 in) steel

plates using a 100 ton (220 kip) hydraulic jack (see Fig. 31 –b). Two

additional LVDTs were inserted in order to measure also the

maximum displacement which, in this case, was not at the loading

points.

1829

P P =19 kips

1829

LEGEND: LVDT'S

LOAD CELLS

610914

CONCRETE STRAIN GAGES

GFRP STRAIN GAGES

914

Load Cell 3

30 ton Hydraulic Jack

Load Cell 2

1 2

914

Load Cell 1

SUPPORTS


a) Test Setup to Determine the Flexural Capacity

1849

P

914

LEGEND: LVDT'S

LOAD CELLS

610914

CONCRETE STRAIN GAGES

GFRP STRAIN GAGES

914

Load Cell 2

Load Cell 1


914

SUPPORTS

914

b) Test Setup to Determine the Shear Capacity

Fig. 31 - Test Setup (all dimensions in mm)

56

3.6 Test Results and Discussion

3.6.1 Failure Modes

For the “flexural” specimen, the first flexural crack was observed on

the longer span when the load was approximately 66 kN (15 kip) on

both spans. As the loads were increased, some of the cracks started to

extend diagonally to form shear cracks (see Fig. 32).

Fig. 32 - Shear Cracks on the Flexure Specimen

The maximum forces, 163 kN (37 kip) and 100 kN (23 kip) on long

and short spans, respectively, represented the maximum load

determining concrete crushing on the top of the longer span,

indicating brittle flexural failure (see Fig. 33).

Shear Crack

57

Fig. 33 - Concrete Crushing in the Compressive Zone

This was immediately followed by a sudden shear failure which also

caused the rupture of the CFRP bars due to kinking (see Fig. 34).

Fig. 34 - CFRP Bars Rupture Due to Kinking

About the “shear” specimen, the first crack was observed at a load

approximately of 89 kN (20 kip) in correspondence of the central

support where the moment was maximum (see Fig. 35). As the load

was increased, the newly formed cracks between the central support

and the loading point on the central span started to extend diagonally

58

to form a shear crack. The failure of the specimen occurred for an

applied load equal to 273 kN (61 kip) due to diagonal tension shear.

Fig. 35 - Failure in the Shear Specimen

3.6.2 Discussion of Test Results

Table 2 compares the experimental results with the theoretical

moment and shear capacities of the slabs computed according to ACI

440 provisions and to the equation developed by Tureyen A. K. and

Frosh R. J. Table 2 – Comparison Experimental Theoretical Results

Experimental Theoretical

Specimen Maximum Bending Moment

kN-m (kip-ft)

Maximum Shear Force

kN (kip)

Moment Capacity

kN-m (kip-ft)

Shear Capacity(1) kN (kip)

Shear Capacity (2) kN (kip)

Shear Capacity(3) kN (kip)

Flexure 101.87 (73.7) 99.7 (22.4) Shear 70.2 (51.8) 142.3 (32.0)

97.9 (72.2)

69.8 (15.7) 97.4 (21.9) 54.7 (12.3)

(1) Computed according to ACI 440 provisions (2) Computed according to Tureyen A. K. and Frosh R. J. (3) Computed with a modified Tureyen A. K. and Frosh R. J. considering c at ultimate

From Table 2 it can be observed that for the flexural specimen the

experimental shear results largely overcame the theoretical shear

capacities of the specimen when calculated according to ACI 440. For

the same specimen the Tureyen A. K. and Frosh R. J. approach

Shear Crack

Central Support

59

overestimated the shear capacity when using the neutral axis at the

service conditions. By removing such assumption and taking c

corresponding to the ultimate conditions of the section, a more

conservative approach can be attained as shown in the last column of

Table 2.

The shear specimen presented a shear capacity larger than the

theoretical, demonstrating again the safe approach of ACI 440. For

this specimen the maximum theoretical bending moment determined

from the Tureyen A. K. and Frosh R. J. approach was only half of the

ultimate moment capacity experimented in the test.

Fig. 36 shows the different trend of measured moment compared to

the theoretical one derived solving the hyperstatic sheme (worked out

in the Test Setup), both related to the central support reaction, R2.

0

10

20

30

40

50

60

70

0 10 20 30 40 50 60 70 80 90

Central Support Reaction [kip]

Mom

ent [

kip-

ft]

M measuredM theoretical

Fig. 36 - Mmeasured Vs. Mtheoretical

60

Fig. 37 shows the load deflection diagram. The change of slope in the

diagram corresponds to the cracking of the specimen.

0

20

40

60

80

100

120

140

160

180

0 10 20 30 40 50 60 70Displacement at the Loading Point [mm]

App

lied

Load

[ kN

]

Fig. 37 – Load deflection Diagram for the Longer Span

of the Flexural Specimen

The same change can be observed, for the same specimen at the same

location, in the diagram of Fig. 38, which shows a comparison

between the experimental and the theoretical moment curvature

diagrams for the “flexure” specimen. The experimental diagram was

determined on the longer span in correspondence of the maximum

bending moment. Compared with the experimental curve, the

theoretical moment curvature diagram showed a lower slope leading

to overestimate the deflection of the bridge.

61

0

20

40

60

80

100

120

140

160

0 5000 10000 15000 20000 25000 30000 35000 40000

Curvature x106 [1/m]

Mom

ent [

kN-m

]

ExperimentalTheoretical

Theoretical Cracking Moment

Theoretical Ultimate Moment

Fig. 38 – Comparison Between Experimental and

Theoretical Moment-Curvature Diagrams

3.7 Conclusions

On the basis of the experimental investigation, it can be concluded

that GFRP bars as passive and CFRP bars as active internal

reinforcement could represent a feasible solution replacing the steel

reinforcement of concrete slab bridges. The prestressing material is

mostly needed for shear purpose rather than for flexure.

According to ACI 440, the shear capacity increases from 30.7 kN (6.9

kip) with mild reinforcement to 69.8 kN (15.7 kip) adding the post-

tensioned reinforcement, while the flexural capacity increases from

76.0 kN-m (55 kip-ft) to 97.9 kN-m (72.2 kip-ft).

Finally, the tools provided by ACI 440 provide a safe design at both

service and ultimate conditions.

62

4 SOUTHVIEW BRIDGE DESIGN

4.1 Introduction

In the following section, the analytical procedures used in the

widening of the Southview Bridge, located in Rolla, Phelps County,

MO, are summarized. The expansion phase included the removal of

the existing curb from the existing reinforced concrete deck to allow

the construction of two new structures adjacent to the original deck in

order to extend the width of the bridge from 3.9 m (13 ft) to 11.9 m

(39 ft). The curb-to-curb width of the resulting bridge is 9.1 m (30 ft).

The two new structures consist of a FRP prestressed/reinforced

concrete deck and a steel reinforced concrete deck as shown in Fig.

39.

WEARING SURFACE5 cm ASPHALT

EXISTING DECKCURB FROM

REMOVE EXISTING

BARRIER WALLCONCRETE

REINFORCED

SIDEWALK (1.5m)NEW CONCRETE

CONCRETE DECKREINFORCED

STEEL

FRP REINFORCED CONCRETE DECK EXISTING REINFORCED CONCRETE

4.3 m1.5 m

12 m2 m3.9 m6 m

FRP REINFORCEDCONCRETE

BARRIER WALL

Fig. 39 - Bridge T0530

The new structure is a box culvert. It consists of a steel reinforced

concrete slab about 25.4 cm (10 in.) thick, as depicted in Fig. 40. The

slab deck is continuous over three intermediate reinforced concrete

vertical walls, and the overall length of the bridge is roughly 12 m (40

ft). The new deck was built on conventional reinforced concrete (RC)

walls. The number of walls is identical to the existing number.

63

Fig. 40 - View of the Existing Bridge

The objective of this section is to provide the structural analysis of the

new FRP concrete bridge deck based on the AASHTO HS20-44

design truck (a detailed analysis is presented in APPENDIX B), and

provide calculations for its design using a combination of non-

traditional corrosion-resistant composites materials.

4.2 Assumptions

The following assumptions are made:

a) Nominal properties for FRP reinforcing material are taken from

the manufacturer published data and considered as initial

guaranteed values to be further reduced to take into account the

environmental reduction factors as given in ACI 440.1R-03

(ACI 440 in the followings);

b) Load configurations are consistent with AASHTO

Specifications;

c) Design carried out according to ACI 440; and

d) Effects due to the skew are neglected.

64

4.3 Structural Analysis

4.3.1 Load Combinations

For the structural analysis of the bridge the definitions of the design

truck and design lane are necessary. This will be addressed in the next

paragraph.

Ultimate values of bending moment and shear force are obtained (in

the following Summary of the Analysis) by multiplying their nominal

values by the dead and live load factors and by the impact factor

according to AASHTO Specifications as shown in Eq. (4.1):

[ ]1.3 1.67( )u d D L Iω β= + + (4.1)

where D is the dead load, L is the live load, βd=1.0 as per AASHTO T

able 3.22.1A, and I is the live load impact calculated as follows: 50 50 0.37 0.30125 10 125

IL

= = = >+ +

(4.2)

and L=3.0 m (10 ft) represents the span length from center to center of

support. The impact factor should not be larger than 0.30, and

therefore the latter value is assumed for the design.

4.3.2 Design Truck and Design Lanes

The analysis of the bridge is carried out for an HS20-44 design truck

load having geometrical characteristics and weight properties shown

in Fig. 41.

65

36 k

N HS20

Variable

4.3 m-9.1 mSpacing

1.8

m

Parapet

30.5

cm

Cle

aran

ce

4.3 m

Parapet

142

kN

142

kN

Fig. 41 - Truck Load on the New Deck

Two loading conditions are required to be checked as depicted in Fig.

42. The HS20-44 design truck load (Fig. 42-a) has a front axle load of

35.6 kN (8.0 kips), second axle load, located 4.3 m (14.0 ft) behind the

drive axle, of 142.3 kN (32.0 kips), and rear axle load also of 142.3 kN

(32.0 kips). The rear axle load is positioned at a variable distance,

ranging between 4.3 m (14.0 ft) and 9.1 m (30.0 ft). Given the specific

bridge geometry, the worst loading scenario is obtained for the

minimum spacing of 4.3 m (14.0 ft) between the two rear axles. The

design lane loading condition consists of a load of 9.3 kN/m (640

lbs/ft), uniformly distributed in the longitudinal direction with

concentrated loads so placed on the span as to produce maximum

stress. The concentrated load and uniform load are considered to be

applied over a 3.0 m (10ft) width on a line normal to the center line of

the lane. The intensity of the concentrated load is represented in Fig.

42-b for both bending moments and shear forces. This load shall be

placed in such positions within the design lane as to produce the

maximum stress in the member.

66

142 kN

4.3 m TO 9.1 m4.3 m

36 kN 142 kN

a) Design Truck (HS20-44)

116 kN FOR SHEAR80 kN FOR MOMENT

9.3 kN/m OVER A 3 m WIDTHTRANSVERSELY UNIFORMLY DISTRIBUTED

b) Design Lane

Fig. 42 - Loading Conditions

4.3.3 Deck Analysis

The deck slab is considered to be a one-way slab system. The width of

the slab, E, to be used in the analysis is provided by AASHTO

(Section 3.24.3.2) as follows: 4 0.06 4 0.06(10 ) 1.4 ( 4.6 )E S ft m ft= + = + = = (4.2)

where S represents the slab length assumed equal to 3.0 m (10 ft).

4.3.4 Flexural and Shear Analysis

Fig. 43 shows a lateral view of the bridge deck when an HS20-44

design truck moves from the right to the left as the value of x1

increases from 0 to L, where L represents the total bridge length.

67

Fig. 43 - Design Truck on the Girder

The values of Pi (i=a,b,c) represent the wheel load as defined by

AASHTO (~18, 71 and 71 kN, namely 4, 16 and 16 kips,

respectively).

Table 3 summarizes values reported in Fig. 43 and reports parameters

used in the calculation of the moment and shear due to dead load. The

analysis and design are carried out for a unit-width strip (~30 cm, 12

in) of slab deck.

Table 3 – Parameters for Girder Analysis

First Span Length, L1 2.89 m (9.49 ft) Second Span Length, L2 3.38 m (11.10 ft) Third Span Length, L3 3.30 m (10.82 ft) Fourth Span Length, L4 2.85 m (9.35 ft) First Load, Pa 17.8 kN (4 kip) Second Load, Pb 71.2 kN (16 kip) Third Load, Pc 71.2 kN (16 kip) Concrete Unit Weight, γc 24 kN/m3 (150 p/f3) Asphalt Unit Weight, γa 17.3 kN/m3 (108 p/f3) Slab Unit-Width, b 30.5 cm (12 in) Slab Height, h 25.4 cm (10 in) Asphalt Thickness, t 15.2 cm (6 in)

As the design truck moves from the right to the left side of the bridge,

fourteen different loading conditions are defined, as shown in Fig. 44.

68

Fig. 44 - Design Truck: Possible Loading Conditions

Three more loading conditions related to the design lane will be

analyzed as reported in Fig. 45. The first loading condition is related

to the maximum positive moment, the second one to the maximum

negative moment, and the third one to the maximum shear.

Fig. 45 - Design Lane: Possible Loading Conditions

69

Fig. 46 shows the moment diagram as the design truck moves on the

bridge following the fourteen loading phases highlighted in Fig. 44.

Fig. 47 shows the moment diagram due to the slab and asphalt layer

self-load. Both diagrams are drawn for a ~30 cm (12 in) strip-width.

0 1.67 3.33 5 6.67 8.33 10 11.67 13.33 1540

30

20

10

0

10

20

30

Axes of the Bridge [m]

Mom

ent [

kN-m

/m]

Fig. 46 - Unfactored Bending Moment Diagrams Due to Live Load

70

0 10 20 30 402

1

0

1

2

Axes of the Bridge [ft]

Mom

ent [

k-ft

/ft]

Fig. 47 - Unfactored Bending Moment Diagrams Due to Dead Load

Ultimate values are obtained by taking into account the maximum

positive and negative moment from Fig. 46 with the load factors

summarized in Eq.(4.1), and by adding the corresponding moment due

to the dead load (Fig. 47) with the load factors taken from the same

equation.

The same diagrams can be drawn for shear as reported in Fig. 48 and

Fig. 49 for live and dead load, respectively, and for a ~30 cm (12 in)

wide unit-strip.

71

0 5 10 1515

10.63

6.25

1.88

2.5

6.88

11.25

15.63

20

Axes of the Bridge [m]

Shea

r For

ce [k

N]

Fig. 48 - Unfactored Shear Diagrams Due to Live Load

0 10 20 30 401.5

1

0.5

0

0.5

1

1.5


Shea

r [ki

ps/ft

]

Fig. 49 - Unfactored Shear Diagrams Due to Dead Load

It has to be underlined that both moment and shear diagrams due to

the dead load have been calculated using a simplified structure where

the distances between supports were assumed to be equal to 3.0 m (10

ft).

72

Similarly, moment and shear diagram related to the design lane

loading condition can be found as depicted in Fig. 50 for the three

structures shown in Fig. 45. The design has been performed for a ~30

cm (12 in) unit strip.

0 5 10 15 20 25 30 35 40 455

4

3

2

1

0

1

2


Mom

ent [

k-ft/

ft]

Maximum Positive Moment

0 5 10 15 20 25 30 35 40 454

3

2

1

0

1

2

3

4


Mom

ent [

k-ft/

ft]

Maximum Negative Moment

73

0 5 10 15 20 25 30 35 40 453

2.5

2

1.5

1

0.5

0

0.5

1


Shea

r [ki

ps/ft

]

Maximum Shear

Fig. 50 - Unfactored Moment and Shear Diagrams

Due to Design Lane Analysis

4.3.5 Summary of the Analysis

Bending moments and shear forces are summarized in Table 4.

Columns (1) and (3) represent the factored coefficient to be applied to

dead and live load, respectively. Columns (2) and (4) show both

unfactored dead and live load moment and shear, respectively, as

taken from the previous figures.

74

Table 4 - Moment and Shear per Unit Strip (Live and Dead Load)

Since the design truck analysis produces the highest stresses on the

slab, only moment and shear related to this analysis will be

considered.

Fig. 51 shows the envelope of the bending moment diagram obtained

when the truck travels on the bridge following all the loading

conditions summarized in Fig. 44.

Fig. 51 - Bending Moment Envelopes (kN-m/m)

Dead Load Live Load Ultimate Loading Conditions

Moment and Shear (1) (2) (3) (4) (5)=(1)(2)+(3)(4)

M+ [kN-m/m]

6.27 (1.38 k-ft/ft)

29.69 (6.55 k-ft/ft)

92.09 (20.3 k-ft/ft)

M- [kN-m/m]

8.69 (1.92 k-ft/ft)

23.52 (5.19 k-ft/ft)

77.56 (17.1 k-ft/ft) Design Truck

Shear [kN /m]

4.92 (1.09 kip/ft)

15.42 (3.40 kip/ft)

160.53 (11.0 kip/ft)

M+ [kN-m/m]

6.27 (1.38 k-ft/ft)

18.21 (4.01 k-ft/ft)

59.41 (13.1 k-ft/ft)

M- [kN-m/m]

8.69 (1.92 k-ft/ft)

17.29 (3.81 k-ft/ft)

59.88 (13.2 k-ft/ft)

Design Lane

Shear [kN /m]

(1.3)

4.92 (1.09 kip/ft)

(1.3)(1.67)(1.3)

12.47 (2.75 kip/ft)

134.25 (9.2 kip/ft)

75

4.3.6 Deflections

The worst loading condition scenario for calculating the slab deck

deflections is reported in Fig. 52. Two concentrated loads simulating

the truck wheels are applied at mid-span of the second and fourth

span. For this analysis, the span lengths have been assumed equal to

30 m (10 ft). To further simplify the analysis, the fourth span of Fig.

52-a could be separately analyzed as depicted in Fig. 52-b. Only the

latter approach is presented here, and the obtained results need to be

considered as an upper bound limit.

a)

b)

∆ LL

PP

P

Fig. 52 - Bridge Deflection Analysis (Live Load)

The deflection due to the dead load (as depicted in Fig. 53) needs to

be added to the live load deflections of Fig. 52.

∆ DL

ωd

l Fig. 53 - Bridge Deflection Analysis (Dead Load)

76

The two settlements can be written as follows: 3

4

8384

0.0065

LLs s

dDL

s s

PE I

E Iω

∆ =

∆ =

l

l (4.3)

where P= 1.3* 71.2 kN (1.3*16 kips) is the wheel load increased by

the impact factor, ωd ~2.6 kN/m (0.179 k/ft) represents the dead load of

the bridge, and Es and Is are the modulus of elasticity of the concrete

and the moment of inertia of the cross-section of the slab,

respectively. The value of the moment of inertia will change

depending on whether the cross-section can be considered cracked or

uncracked.

4.4 Design

4.4.1 Assumptions

The design of the internal FRP reinforcement is carried out according

to the principles of ACI 440. The properties of concrete, steel and FRP

bars used in the design are summarized in Table 5. The reported FRP

properties are guaranteed values.

77

Table 5 - Material Properties

Concrete Compressive

Strength f`c (MPa)

FRP Internal

Reinforcement Type

FRP Bar Size

FRP Tensile Strength f*

fu (MPa)

FRP Tensile Strain ε*

fu

FRP Modulus

of Elasticity Ef (GPa)

f9 (#3) 758 (110 ksi) 0.018 40.8

(5,920 ksi)

f13 (#4) 689 (100 ksi) 0.017 40.8

(5,920 ksi) f19 (#6) 621

(90 ksi) 0.015 40.8 (5,920 ksi)

GFRP

f22 (#7) 586 (85 ksi) 0.014 40.8

(5,920 ksi)

41.4 (6,000 psi)

CFRP f19 (#3) 2068 (300 ksi) 0.017 124.1

(18,000 ksi)

Material properties of the FRP reinforcement reported by

manufacturers, such as the ultimate tensile strength, typically do not

consider long-term exposure to environmental conditions, and should

be considered as initial properties. FRP properties to be used in all

design equations are given as follows (ACI 440): *

*

fu E fu

fu E fu

f C f

Cε ε

=

= (4.4)

where ffu and εfu are the FRP design tensile strength and ultimate strain

considering the environmental reduction factor (CE) as given in Table

7.1 (ACI 440), and f*fu and ε*

fu represent the FRP guaranteed tensile

strength and ultimate strain as reported by the manufacturer (see Table

5). The FRP design modulus of elasticity is the average value as

reported by the manufacturer.

78

4.4.2 Slab Design

4.4.2.1 Flexural Design

The flexural design of a FRP reinforced concrete member is similar to

the design of a steel reinforced concrete member. The main difference

is that both concrete crushing and FRP rupture are potential

mechanisms of failure. As an FRP reinforced concrete member is

usually less ductile than the correspondent steel reinforced concrete

member, the strength reduction factor, φ, needs to be revisited

according to Eq. (4.5) (ACI 440):

0.50

1.420.70 1.4

f fb

ffb f fb

fb

f fb

if

if

if

ρ ρ

ρφ ρ ρ ρ

ρ

ρ ρ

≤= < < ≥

(4.5)

where ρf is the FRP reinforcement ratio and ρfb represents the FRP

reinforcement ratio producing balanced failure condition.

Fig. 54 shows the trend of the strength reduction factor as a function

of both concrete compressive strength and the number of GFRP bars

used.

79

2000 4000 6000 80000.4

0.5

0.6

0.7

0.8

#[email protected]"f`c (psi)

Phi F

acto

r as p

er A

CI 4

40

F19@15 cm 19@12 cm [email protected] cm F

Fig. 54 - Strength Reduction Factor

By increasing the concrete compressive strength, the value of the φ

factor changes from 0.7 to 0.5 as the failure mode moves from

concrete crushing to FRP rupture.

Fig. 55 shows the nominal and factored flexural capacities of the

bridge deck as a function of the concrete compressive strength and the

number of FRP bars installed (same legend of Fig. 54).

2000 4000 6000 800020

30

40

50

60

#[email protected]"f'c [psi]

Mn

[k-ft

/ft]

2000 4000 6000 8000

15

20

25

30

35

#[email protected]"f'c [psi]

phiM

n [k

-ft/f

t]

a) Nominal b) Factored Fig. 55 - Flexural Capacities of the Bridge

Table 6 summarizes the properties and the flexural capacity of the

bridge deck corresponding to a cross section with a f19 (#6) Aslan

80

100 GFRP as mild reinforcement at ~15 cm (6 in) center-to-center.

Prestressing FRP tendons made out of f9 (#3) Aslan 200 CFRP bars

are installed at ~23 cm (9 in) center-to-center and post-tensioned after

the concrete deck is cured. Calculations are carried out for a ~30 cm

(12 in) unit strip.

Table 6 - Slab Geometrical Properties and Internal FRP Reinforcement

Overall Height

of the Slab,

h

Area of Tension GFRP Reinforcement,

Af

GFRP Effective Depth,

d

Area of Prestressed

CFRP Tendons,

Ap

CFRP Effective Depth,

dp

Flexural Capacity

φMn

[cm] [cm2/m] [cm] [cm2/m] [cm] [kN-m/m] 25.4

(10 in) 19.39

(0.916 in2/ft) 20.6

(8.125 in) 2.13

(0.101 in2/ft) 17.8

(7.0 in) 92.8

(20.5 k-ft/ft)

The above flexural capacity is related to both positive and negative

moment regions, and it is larger than the required demand previously

shown in Table 4.

The prestress strain in the tendons is equal to 65% of their ultimate

strain. All CFRP losses are assumed to be 30% of the initial

prestressing strain.

4.4.2.2 Shear Design

The shear capacity of FRP reinforced concrete sections is calculated

following the principles of ACI 440. Particularly, the concrete

contribution to the shear capacity, Vc,f, can be expressed as follows:

,f f

c f cs s

A EV V

A E= (4.6)

81

where the ratio (AfEf/AsEs) takes into account the axial stiffness of the

FRP reinforcement as compared to that of steel reinforcement, and Vc

is given as follows (ACI 318-99):

'(0.6 700 )u pc c p

u

V dV f bd

M= + (4.7)

As reported in Eq. (4.6) can be found by determining the area of steel

reinforcement required to match the factored FRP flexural capacity,

φMn. Table 7 summarizes the steel and GFRP design and the assumed

value of As.

Table 7 - Steel and GFRP Flexural Design of the Deck (No Prestress Used)

Description Steel Design GFRP Design Deck Slab, h 25.4 cm (10 in) 25.4 (10 in) Deck Width, b 30.5 cm (12 in) 30.5 (12 in) Effective Depth, d 20.6 cm (8.125 in) 20.6 (8.125 in)

Reinforcement Area f13@25cm=11.2cm2/m (#4@10”=0.528 in2/ft)

f19@11cm=25.3 cm2/m (#[email protected]”=1.195 in2/ft)

Factored Flexural Capacity,φMn

94.3 kN-m/m (20.8 k-ft/ft)

93.0 kN-m/m (20.5 k-ft/ft)

The concrete contribution to the shear capacity of the member yields:

11.0(7 /12)0.6 6000 700 (12)(7) 379.4 / ( 26.0 / )17.1cV kN m kips ft = + = =

(4.8)

and from Eq. (4.6):

,(1.195)(5,920) 26.0 175 / ( 12.0 / )

(0.528)(29,000)c fV kN m kips ft= = = (4.9)

Finally, the factored shear capacity is φVn=φVc,f=0.85*175=148.9

kN/m (=10.2 kips/ft). This value is slightly smaller than the shear

demand Vu=160.5 kN/m (=11.0 kips/ft, see Table 4). The value can be

82

accepted because the analysis has been performed on the centerline of

the support, while it is allowed to evaluate both Vu and Mu at a cross

section flush with the vertical wall representing the support. In this

case, Vu will not change substantially, while Mu will have an

appreciable lower value. Therefore (while keeping Vudp/Mu<1), Vc

expressed by Eq. (4.8) will increase.

Consistent with section 3 it is to be noted that the prestressing force is

most needed for shear purpose rather than flexure. In fact, if the

prestressing action was not considered, the concrete contribution to

the shear capacity would have been Vc=220.4 kN/m (=15.1 kips/ft),

and the final factored shear capacity of the bridge would have been

equal to φVn=86.09 kN/m << 148.9 kN/m (φVn=5.9 kips/ft << 10.2

kips/ft); therefore the post-tensioning allowed to increase the shear

capacity of the slab more than 40%.

As an alternative approach to the shear capacity of the bridge deck,

the following equation (Tureyen A. K. and Frosh R. J., 2003), now

under consideration for adoption by ACI Committee 440, could be

used:

`, 5c f cV f bc= (4.10)

where c is the position of the neutral axis at service. This approach is

justified by the parametric analysis laid out in the next Fig. 56.

83

Fig. 56 - Rationale for Proposed Shear Strength

The values of c can be determinate using the approach shown in

paragraph 4.4.3.1 afterwards:

5 6000 12 3.295= 68.10kN(=15.31 kip)cV = ⋅ ⋅ ⋅ (4.11)

Finally, the factored shear capacity is φVn=φVc,f=0.85(213.2) =189.9

kN/m (0.85(14.61)=13.01 kip/ft) larger than Vu=160.5 kN/m (11.0

kip/ft).

84

4.4.2.3 Temperature and Shrinkage Reinforcement

GFRP reinforcement perpendicular to the main flexural reinforcement

is required to control both crack width and shrinkage of the concrete.

The equation adopted by ACI 440 can be written as follows:

,60,0000.0018 0.0036s

f tsfu f

Ef E

ρ = ≤ (4.12)

where ffu (psi) is defined in Eq.(4.4), and Es an Ef are the elastic

moduli of steel and GFRP, respectively. The area of GFRP

reinforcement deemed necessary for temperature and shrinkage can be

expressed as follows:

, ,f ts f tsA bhρ= (4.13)

and it is subdivided in two layers, each close to one of the concrete

surfaces. In the previous equation, b and h represent unit width and

height of the cross-section, respectively (b=30.5 cm, 12 in, and

h=25.4 cm, 10 in).

It is suggested to use a f13 (#4) Aslan 100 GFRP bar spaced at ~30

cm (12 in) center-to-center, as depicted in Fig. 57 representing a cross

sectional view of the bridge deck.

Fig. 58 shows a longitudinal view of the prestressing CFRP tendons.

85

F

F

19 ASLAN 100 GFRP AT 15 cm ON CENTERS (G1)

9 ASLAN 200 CFRP (C1)PRESTRESSING TENDONSSPACED 23 cm ON CENTERS

TEMP. & SHRINK.13 ASLAN 100 GFRP

AT 30 cm ON CENTERS

19 ASLAN 100 GFRPAT 15 cm ON CENTERS

(G2)

(G3) AND (G4)

9 ASLAN 100 GFRPCHAIRS (G5)

FF

F

Fig. 57 - Bridge Internal FRP Reinforcement: Section at Mid-Span

Fig. 58 - CFRP Prestressing Tendons

a) Longitudinal View

F9 ASLAN 200 CFRPPRESTRESSING TENDONSSPACED 23 cm ON CENTERS

A

A

b) Detail

86

4.4.3 Serviceability

Unlike steel reinforced concrete sections, members reinforced with

FRP bars have relatively small stiffness after cracking. Therefore,

serviceability requirements like crack width and long-term deflection

need to be specifically tailored for composite structures as highlighted

in ACI 440. In the following two sections, both crack width and long-

term deflection checks will be presented.

4.4.3.1 Crack Width

The service moment per unit strip of slab deck can be calculated

starting from the data of Table 4 as follows:

1.3 1.381 (1.3)(6.549) 44 / ( 9.9 / )s DL LLM M M kN m m k ft ft= + = + = − = − (4.14)

where MDL and MLL represent moment due to dead and live load,

respectively.

Crack width of FRP reinforced flexural members can be expressed as

suggested by ACI 440 as follows:

32200b f c

f

w k f d AE

β= (4.15)

where β is the ratio of the distance from the neutral axis to extreme

tension fibre to the distance from the neutral axis to the center of

tensile reinforcement, kb is a bond-dependant coefficient equal to 1.2,

ff represents the stress at service in the FRP, dc is the thickness of the

concrete cover measured from extreme tension fiber to the center of

the bar, A is the effective tension area of concrete defined as the area

87

of concrete having the same centroid as that of tensile reinforcement,

divided by the number of bars (N=2), and Ef is the modulus of

elasticity of FRP. The above mentioned terms, can be written as

follows:

1

1

2c

c

h cd c

d h dd bAN

ββ

β−

=−

= −

=

(4.16)

Assuming a cracked concrete cross-section at service as shown in Fig.

59, the stress in the GFRP, ff =Ef ef, can be calculated by solving the

following system of equations:

c c p p p f f f pi

c c p p p p f f f s pi

pc

p

c f

1 E bc A E A E N21 h c h hE bc A E d A E d M M2 2 3 2 2

c d c

c d c

ε ε ε

ε ε ε

εε

ε ε

− − =

− + − + − = − = − = −

(4.17)

where Mpi and Npi represent the axial load and the bending moment to

the centroid of the section, induced by the post-tensioning of the

CFRP bars.

By solving Eq. (4.17) and finding the unknown fε , the following

value for the stress in the GFRP can be evaluated as follows, after :

f f ff E 13.58MPa( 1.97 ksi)ε= = = (4.18)

88

Fig. 59 - Stress and Strain at Service

The crack width calculated with Eq. (4.15) yields w=0.086mm

(0.0034 in) which is smaller than the allowed values suggested by ACI

440 and equal to 0.51mm (0.02 in).

4.4.3.2 Long-Term Deflections

Based on the conservative deflection analysis carried out on Section

B.3.3, the long-term deflection can be calculated as suggested by ACI

440 as follows:

( 0.2 )LL DL LLλ∆ = ∆ + ∆ + ∆ (4.19)

where λ=1.2 represents the multiplier for additional long-term

deflection as recommended in ACI 440. Assuming the concrete cross-

section uncracked because of the presence of the prestressing tendons,

both concrete modulus of elasticity and moment of inertia can be

expressed as follows: '

3

57000

12

cE f

bhI

=

= (4.20)

where b~30 cm (=12 in) for dead load analysis, and b=E=1.4 m (4.6

ft) for live load analysis. Eq. (4.19) yields to ∆=1.30 mm (0.051 in)

89

smaller than the suggested AASHTO value of l/800=3.81mm (0.15

in).

4.4.3.3 Slab Creep Rupture and Fatigue

To avoid creep rupture of the FRP reinforcement under sustained

loads, the stress level in the FRP bar should be limited to the value

suggested in ACI 440. Specifically, when GFRP reinforcement is

used, the stress limit has been set to be equal to 0.20 ffu ~87 MPa (12.6

ksi). The stress at service in the FRP can be found as:

(9.9)(12) 114.5 ( 16.6 )0.9430.916 8.125

3 3

sf

f

Mf MPa ksicA d

= = = = − −

(4.21)

Once again, because the bridge is mostly uncracked at service due to

the prestress, the above findings are conservatives, and the value

obtained from Eq. (4.21) can be considered acceptable.

4.5 Load Rating

Bridge load rating calculations provide a basis for determining the

safe load carrying capacity of a bridge. According to the Missouri

Department of Transportation (MoDOT), anytime a bridge is built,

rehabilitated, or reevaluated for any reason, inventory and operating

ratings are required using the Load Factor rating. All bridges should

be rated at two load levels, the maximum load level called the

Operating Rating and a lower load level called the Inventory Rating.

The Operating Rating is the maximum permissible load that should be

allowed on the bridge. Exceeding this level could damage the bridge.

90

The Inventory Rating is the load level the bridge can carry on a daily

basis without damaging the bridge.

The Operating Rating is based on the appropriate ultimate capacity

using current AASHTO specifications (AASHTO, 1996). The

Inventory Rating is taken as 60% of the Operating Rating.

The vehicle used for the live load calculations in the Load Factor

Method is the HS20 truck. If the stress levels produced by this vehicle

configuration are exceeded, load posting may be required.

The tables below show the Rating Factor and Load Rating for this

bridge. The method for determining the rating factor is that outlined

by AASHTO in the Manual for Condition Evaluation of Bridges

(AASHTO, 1994). Equation (4.22) was used:

( )1

2 1C A DRF

A L I−

=+

(4.22)

where: RF is the Rating Factor, C is the capacity of the member, D is

the dead load effect on the member, L is the live load effect on the

member, I is the impact factor to be used with the live load effect, A1

is the factor for dead loads, and A2 is the factor for live loads. Since

the load factor method is being used, A1 is taken as 1.3 and A2 varies

depending on the desired rating level. For Inventory rating, A2 = 2.17,

and for Operating Rating, A2 = 1.3.

To determine the rating (RT) of the bridge Equation (4.23) was used:

( )RT RF W= (4.23)

In the above equation, W is the weight of the nominal truck used to

determine the live load effect.

91

For the Southview Bridge, the Load Rating was calculated for a

number of different trucks, HS20, H20, 3S2, and MO5. The different

ratings are used for different purposes by the bridge owner. For each

of the different loading conditions, the maximum shear and maximum

moment were calculated. Impact factors are also taken into account

for Load Ratings. This value is 30% for the Southview Bridge. The

shear and moment values for the deck are shown in below in Table 8.

Table 8 - Maximum Shear and Moment due to Live Load

Truck Maximum Shear (kN)

Maximum Moment (kN-m)

Maximum Shear with

Impact (kN)

Maximum Moment

with Impact (kN-m)

HS20 15.1 (3.40 kips)

9.06 (6.55 k-ft)

19.66 (4.42 kips)

11.78 (8.52 k-ft)

MO5 14.7 (3.30 kips)

6.47 (4.68 k-ft)

19.66 (4.29 kips)

8.41 (6.08 k-ft)

H20 12.7 (2.86 kips)

5.93 (4.29 k-ft)

14.55 (3.72 kips)

7.70 (5.57 k-ft)

3S2 13.0 (2.93 kips)

5.89 (4.26 k-ft)

16.95 (3.81 kips)

7.66 (5.54 k-ft)

Table 9 below gives the results of the Load Rating pertaining to

moment and Table 10 shows the results for shear. All calculations for

the load rating are located in APPENDIX B.

Table 9 - Rating Factor for the New Slab (Bending Moment)

Truck Rating Factor (RF)

Rating (RT) (Tons)

Rating Type

HS20 1.793 64.5 Operating HS20 1.074 38.7 Inventory MO5 2.482 89.4 Operating H20 2.330 46.6 Posting 3S2 2.345 85.9 Posting

92

Table 10 - Rating Factor for the New Slab (Shear)

Truck Rating Factor (RF)

Rating (RT) (Tons)

Rating Type

HS20 2.152 77.5 Operating HS20 1.289 46.4 Inventory MO5 2.217 81.2 Operating H20 2.202 44.0 Posting 3S2 2.148 78.7 Posting

Since the factors RF are greater than 1 then the bridge does not need

to be load posted. In addition, from Table 9 and Table 10 the

maximum operating and inventory load can be found as 64.5T and

38.7T respectively.

4.6 Southview Bridge Drawings

The following drawings show the main characteristics of the

Southview bridge deck, consistent with the previously tested

specimens and with the design presented in this section.

Drawing 1 and Drawing 2 show a plan strip view with the transverse

and longitudinal section of the deck, respectively. Drawing 3 details

profile, section, top and bottom view of the slab; FRP reinforcement

details at the midspan and at the supports are also shown.

93

plastic ductwork

safety ductwork

SOUTHVIEW DRIVE BRIDGEFRP post-tensioned slab

19 ASLAN 100 GFRP (G1,G2)

13 ASLAN 100 GFRP(G3,G4)

A A

Plan Strip15

91

30

30

23

safety ductwork

1.5 m long bolsters

CHAIRS (G5)140 9 ASLAN 100 GFRP

AT 15 cm ON CENTERS

Section A-A40 19 ASLAN 100 GFRP (G1)

AT 30.5 cm ON CENTERS

SPACED 23 cm ON CENTERS


PRESTRESSING TENDONS26 9 ASLAN 200 CFRP (C1)

9 ASLAN 200 CFRP (C1)F

F

F

F

F

F

AT 15 cm ON CENTERS

F

40 19 ASLAN 100 GFRP (G1)F

Drawing 1 - Section A-A

Plan Strip

12.420.41 0.41

0.412.853.303.382.890.41

9 ASLAN 200 CFRP (C1)

622

F

F

F

F

F43 13 ASLAN 100 GFRP AT 30.5 cm ON CENTERS (G4)

622

43 13 ASLAN 100 GFRP AT 30.5 cm ON CENTERS (G3)


1.5 m LONG BOLSTERS AT 60 cm

CHAIRS AT 3'140 9 ASLAN 100 GFRP (G5)

section B-B

25

BB

Drawing 2 - Section B-B

94

6.35

21.6421.64

6.31

13.23

0.41 0.41

0.410.41

Bottom chairs

UP CHAIRS (G5)

3.00

0.61

0.038

0.178

0.038

0.254

chairs detail

22.86

15.24

0.254

0.038

0.178

0.0380.038

0.177

1.22

0.254

section 2-2

Bottom chairsbottom chairs

13 GFRP (G4)

13 GFRP (G3)

n.8 19 GFRP (G1)

n.8 19 GFRP (G2)

n.5 9 CFRP (C1)

section 1-1

up chairs (G5) up chairs (G5)

safety ductwork

45.7245.72

safety ductwork

wal

l axis

wal

l axis

plastic ductwork

plastic ductwork

plastic ductwork

0.06

0.25

1.22

22.86

15.24

0.61

0.039

2

2

1

1

n.22 13 GFRP (G4)n.21 13 GFRP (G3)

n.6 19 GFRP (G2)n.4 9 CFRP (C1)

Bottom reinforcment plan Top reinforcement plann.6 19 GFRP (G1)n.4 9 CFRP (C1)

PLAN - 3 feet slab's strip


(L=13.7 m)

(L=13.1 m)

(L=13.1 m)

DETAIL - FRP Reinforcement

9 ASLAN 200 CFRP PRESTRESSING TENDONS AT 23 cm ON CENTERS (C1)

wal

l axis

SECTION - deck's FRP reinforcement




PROFILE - Southview Drive Bridge

2.853.303.382.89

12.42

F

F

F

F

F

F F

F F

F F

F

F

F

F

F

Fn.8 19 GFRP (G1)

13 GFRP (G3)F

F 13 GFRP (G4)

n.8 19 GFRP (G2)F

n.5 9 CFRP (C1)F Drawing 3 - Ultimate Design

95

5 SOUTHVIEW BRIDGE DECK INSTALLATION

5.1 Introduction

This section details the installation of Southview bridge deck,

conducted by Master Contractors LLC, overseen and documented by

the author as part of this thesis.

As outlined in the first section, the slab is 25 cm (10 in) thick, 12 m

(40 ft) long and 6 m (19.83 ft) wide. It is supported by three

intermediate reinforced concrete vertical walls. The three span lengths

are, from North to South, 2.89 m (9.49 ft), 3.38 m (11.10 ft) and 3.30

m (10.82 ft), respectively, on centers.

The material and construction specifications are detailed in Appendix

C.

5.2 Pre-construction Meetings

In order to ensure a successful completion of the project, extensive

planning and collaboration was required between the different entities

involved in the construction: the City of Rolla in charge of the

construction of walls and abutments expansion and of the post-

tensioning system which was optimized for the particular field

application as part of a cooperation between University of Missouri

Rolla (UMR) and the Rolla Technical Institute (RTI) in the person of

Mr. Max Vath.

A first meeting was held on Tuesday, July 27th, 2004 at the City of

Rolla. This meeting was attended by representatives of the City: the

96

engineer, David Brown and the foreman construction, Mr. Bill

Cochran; the contractor in charge of the construction of the slab, Mr.

Jason Cox from Master Contractors LLC, and Dr. Antonio Nanni and

the author from UMR.

The three main topics were discussed at this meeting concerning

issues of constructability, materials, and structural concerns.

• Constructability issues dealt with the schedule of construction

and access to the bridge deck during construction. The existing

lane of the bridge was the only access that could be used to

reach the construction site; therefore carefulness had to be paid

to the trucks that continuously passed on the bridge in order to

create a safe environment for the workers.

• Material issues dealt with quality assurance testing: the

University of Missouri-Rolla volunteered testing concrete

samples extracted during the pouring and the steel rebar.

• Structural issues dealt on the discussion of details such us

laying GFRP bars on the central wall in order to create a

connection between slab and walls, the walls had to have

different height due to the presence of Neoprene pads only on

two of the three walls, and the construction of the formwork

supporting the slab by the City of Rolla.

The second pre-construction meeting was held at the construction site

on Thursday, August 12th 2004. The meeting was attended by the

same people present at the first one, with the addition of one the

bridge deck designers, Dr. Nestore Galati from UMR.

The main topics pertained to concrete, formwork and barrier:

97

• Concrete issues dealt with quality and required characteristics

of concrete to be used for the construction of the deck. A high

slump value was determined to be necessary and it had to be

achieved with the addition of super-plasticizers in order to make

sure the matching of an adequate strength in a short period time.

The City of Rolla was in charge of providing the concrete.

• Formwork: In order to facilitate the post-tensioning a formwork

longer than the slab itself had to be built on both sides of the

bridge (see Fig. 87). This exceeding part of the formwork would

have been cut after the post-tensioning operations, thus not

affecting the construction of two more box cells, scheduled after

the installation of the deck.

• Barrier: During the construction of the walls it was also decided

to build the barrier on the “FRP side” of the bridge using GFRP

rebars instead of steel rebars.

Such decision would in-fact allow the comparison between the

durability of the FRP and the steel barrier to be built on the

opposite side of the bridge, as already discussed and shown

before.

5.3 Substructure Construction

5.3.1 Footing and Floor Construction

As highlighted before, the erection of the substructure and the

extension of the existing abutments and walls were performed by the

City of Rolla employees prior to the slab construction.

98

The 21st of July the works started with the removal of the guardrail,

the wetting of the basement and the digging of the weed on the “FRP

side”, as shown in Fig. 60.

Fig. 60 - Wetting of the Basement and Digging of the Weed

The following days, after the excavation of the bridge site and the

moving of soil, the suction of the water overflowing from the creek

was the main operation to deal with; indeed the area where the footing

had to be built (see Fig. 61) was often and very easily filled with water

either due to rain or to a mizzle. In order to remove the stagnant water

different devices were utilized, such as electric pumps or a pipe

allowing the water to flow from the creek getting over the basin to the

opposite side of the stream (see Fig. 62).

99

Fig. 61 - Work Area after the Water Filling

Fig. 62 - Devices to Remove or Avoid the Water

On July 27th, the formwork and the steel reinforcement for the footing

were placed (see Fig. 63), then the following day, after removing the

remaining water, the footing was poured (see Fig. 64). The concrete

slump was found to be 10 cm (4in).

100

Fig. 63 - Footing Formwork

Fig. 64 - Pouring of the Footing

Fig. 65 shows the Slump Test and the casting of concrete cylinders

used to determine the concrete strength. The footing was built by the

end of the working day (see Fig. 66).

101

Fig. 65 - Slump Test and Concrete Cylinders Casting

Fig. 66 - Footing after Pouring

On July 29th the floor reinforcement was placed, after filling the voids

between the footing beams with gravel (see Fig. 67).

Fig. 67 - Floor Reinforcement

102

As highlighted before several delays were caused by the bad weather

conditions. Fig. 68 shows the site conditions after a thunderstorm. The

casting of the floor was completed on the 3rd of August.

Fig. 68 - Water Overflowing and Casting of the Floor

5.3.2 Abutments Construction

The abutments construction started after replacing the former

abutments caps, since the existing ones were in very bad conditions.

Then, the formwork and the reinforcement of the abutments were

placed quickly (see Fig. 69). Fig. 70 shows the device used to tie the

steel reinforcement together.

August 5th the new abutments were cast (see Fig. 71).

103

Fig. 69 - Laying of Abutments Reinforcement and Formwork

Fig. 70 - Use of the Steel Ties Gun

Fig. 71 - Casting of the New Abutments

104

5.3.3 Walls Construction

Prior to the walls construction, during the curing of the new

abutments, the dowels for the walls were fixed on the existing floor;

thus the first wall reinforcement and formwork were placed (as shown

in Fig. 72).

The following Monday, August 9th, the first wall was cast (see Fig.

73).

Fig. 72 - Reinforcement and Formwork of the First Wall

Fig. 73 - First Wall

The second wall to be built was the one close to the other abutment. In

this way it was possible to reuse the first wall formwork. The central

105

wall formwork had to be lower than the others to allow the insertion

of the GFRP bars anchoring the slab to that wall.

The construction of the central wall was complicated by the lack of

room and the short distance between vertical steel rebars to which the

GFRP rebars had to be tied (see Fig. 74 and Fig. 75). In addition a

central groove on the top of the wall was made to strengthen the

connection between wall and slab (see Fig. 76).

The prefatory part of the project was completed on August the 12th

(see Fig. 77 and Fig. 78).

F

3.8

76.2

19.7

0.950.9538.1

n.40 9 ASLAN 100 GFRP

76.2

10.25.1

38.1

76.2

1.95.50

3.81.9

25.4

20.725.4

5.1

10.2

7.67.6

Fig. 74 - GFRP Anchoring Detail

Fig. 75 - Laying of the GFRP Rebars

106

Fig. 76 - Making of the Notch

Fig. 77 - Central Wall

Fig. 78 - Completion of the Prefatory Part of Southview Project

107

5.4 Slab Construction

Prior to the construction of the slab all the material needed was

computed. The bill of materials and equipment used for the

installation of the slab is reported in the following section.

5.4.1 Bill of Material

Table 11 summarizes the amount of FRP reinforcement needed.

Table 11 - Bill of FRP Material

No. Size Length Mark Location Fiber Type

Bending Sketches

26 f9 #3

16.15 m (53’-0”) C1 Deck Carbon

40 f19 #6

13.11 m (43’-0”) G1 Deck Glass

40 f19 #6

13.11 m (43’-0”) G2 Deck Glass

43 f13 #4

5.79 m (19’-0”) G3 Deck Glass

43 f13 #4

5.79 m (19’-0”) G4 Deck Glass

140 f9 #3

0.97 m (3’-2”) G5 Deck Glass

43 f19 #6

1.95 m (6’ -1”) G6 Barrier Glass

43 f13 #4

1.85 m (5’ -9”) G7 Barrier Glass

10 f13 #4

13.11 m (43’-0”) G8 Barrier Glass

40 f19 #6

1.70 m (5’ -7”) G9 Deck Glass

40 f9 #3

1.24 m (3’-9”) G10 Deck-Wall Glass

Moreover the following materials were used:

1. Neoprene pads: NEWLON 60 durometer molded neoprene

for load-bearing applications. Considering ~15x10 cm2 (6x4

108

in2) pads, 1.3 cm (1/2 in) thick, ~15 cm (6 in) from each other,

each wall needed 24 pads. Since only two of the three walls

needed the pads and accounting also for the two abutments, a

total of 96 pads were needed (~1.5 m2, about 2300 in2).

2. Styrofoam: ~6.6 m2 (71 ft2) of styrofoam were required to

place around the pads.

3. Chairs: ~1.5 m (5 ft) long bolsters at ~60 cm (2 ft) spaced

were deemed necessary to support the bottom longitudinal

GFRP bars, considering ~6 m (20 ft) of width and ~12 m (40

ft) of length of the slab, 4 sets of 20 chairs, namely 80 chairs

were ordered.

4. Ties: 5000 plastic zip ties were used to tie the bars and the

duct, in order to have a complete “steel free” structure.

5. Plastic duct: The bridge is ~13 m (43 ft) long, so a ~13.7 m

(45 ft) long duct was necessary for each tendon, plus three

~1.5 m (5 feet) long pieces of duct coming out from the deck

and inserted through a T connector to the previous duct in the

way to inject the grout. Hence the needed amount was about

~18 m (60 ft) of ~2.3 cm (9/10 in) diameter plastic duct for

each tendon for a total amount of ~400 m (1300 ft). Adding

thirteen ~13.7 m (45 ft) long safety duct (plus two ~1.5 m long

coming out pieces) plus a supply, the overall duct length was

~700m (2300 ft).

6. T connectors: 3 T connectors of ~2.5 cm (1 in) internal

diameter for each tendon trace were provided, so as to create a

straight duct with a 3rd piece coming out in the way to inject

the grout. 26 tendons request 78 T connectors, plus 26 T

109

connectors for the safety duct is 104 T connectors, with

supplies, hence 110 T connectors overall.

7. Injection grout: The Sikadur 300 PT is the grout for injection

that was used for the bridge. Every bag is about ½ cubic foot,

thus considering 39 per ~13.7 m (45 ft) long duct ~2.3 cm

(9/10 in) diameter with a f9 (#3) tendon inside, ~0.26 m3 (9.35

ft3) of grout was ordered, and namely 20 bags of Sikadur 300

PT were provided.

8. Concrete: The ~12*6*0.25 m3 (40*20*0.83 ft3) slab required

~21 m3 (750 ft3) of concrete, with supplies.

In addition the following tools were used to pull the CFRP tendons:

1. Stereophone packs: two ~37*25*25 cm3 (12*10*10 in3)

stereophone packs were assembled in order to house the

tendons, at the same time providing enough room to push the

wedges inside the chucks before cutting the post-tensioned

bars. More details will be itemized afterwards.

2. Chucks: Pulling 13 tendons per time, 2 chucks for each

tendon and two more after the two hydraulic jacks when the

load was applied were deemed needful. Therefore 28 chucks

were requested overall. Each chuck includes an outer steel

cylinder, a four-piece steel wedge and an inner copper sleeve

(The same type used for the specimens load test, see section

3).

3. Steel Plates: Each tendon needed two steel plates between the

edges of the slab and the pulling machine. Thus twentysix

~(10*10*1.3) cm3 (4*4*1/2 in3) steel plates were required. All

the steel plates had a 1.9 cm (0.75 in) inner hole, in order to

avoid excessive transversal movement of the tendons.

110

Additionally the following equipment was used for the prestressing

operations:

• 2 hydraulic jacks (40 tons), to pull the tendons from each side

of the slab.

• 2 load cells (~222 kN, i.e. 50 kips each), to monitor the applied

load.

• 1 injection pump, to inject the grout inside the duct.

• 4 wedges hammers to push the wedges inside the chucks.

• Orange box, an easily transportable system that records the

load, the strains and the deflections.

• Wooden Boards, to build a surface suitable for the post-

tensioning, given the skewness of the slab.

5.4.2 Outline of tasks

The installation of the deck proceeded based on the following tasks:

• Partial embedding of anchoring GFRP bars into the central wall,

already described before.

• Laying of neoprene pads on the two external walls and on the

abutments.

• Setting of the formwork.

• Installation of bottom chairs, bottom longitudinal and

transversal GFRP bars.

• Installation of top chairs, top longitudinal and transversal GFRP

bars.

• Placing of the GFRP reinforcement for the barrier.

• Placing of ductwork, T connectors and CFRP bars inside the

ductwork; setting of the safety ductwork.

111

• Pouring and curing of concrete.

• Post-tensioning of CFRP tendons.

• Injection of grout inside the ductwork.

• Cutting of the tendons outside of the slab after the curing of the

grout.

• Cutting of the edges of the slab.

5.4.3 Investigation of the Slab Construction

On August 13th the installation work started, but, given the lack of

some hangers (see details in Fig. 79 and Fig. 80), only on August 17th

the activities could restart.

Fig. 79 - Formwork Hangers Details

112

Fig. 80 - Laying of the Formwork

After the installation of the formwork, on August the 18th the

following operation was the gluing of the Styrofoam on the two

abutments and on the two walls that didn’t have the GFRP anchoring

rebars (See Fig. 81).

Fig. 81 - Gluing and Placing of Styrofoam

Hence the Neoprene pads were placed in the room created in the

Styrofoam as shown in Fig. 82. The neoprene pads were used to avoid

113

restraining horizontally the slab and therefore to effectively post-

tension it.

Fig. 82 also shows the plastic chairs that supported the bottom layer of

GFRP mild reinforcement.

Fig. 82 - Neoprene Pads and Plastic Chairs Details

The placing of the FRP reinforcement was speed up thanks to the

presence of many students from RTI, leaded by Mr. Harold Martin,

who envisioned this as an opportunity to teach to his students new

technologies (see Fig. 83 and Fig. 84).

Fig. 83 - Laying of Bottom Layer of GFRP Rebars

114

Fig. 84-The RTI Team with (from down on the left) Mr. Cochran,

the Author and Mr. Cox

On August 21st the top layer was also laid after placing all the GFRP

top chairs. Furthermore, according to the design reported in Fig. 85,

also the GFRP rebars for the barrier were placed as shown in Fig. 86:

TEMP. & SHRINK.13 ASLAN 100 GFRP

AT 25 cm ON CENTERS3.8

20

AT 30.5 cm ON CENTERS

AT 25 cm ON CENTERSTEMP. & SHRINK.

(G8)

(G7)

(G6)

(G8)

Bridge Deck

F13 ASLAN 100 GFRP

19 ASLAN 100 GFRPF

F13 ASLAN 100 GFRPF

25.4

111.8

POST-TENSIONED FRP DECK

AT 30 cm ON CENTERS

SIDEWALK

Fig. 85 - Barrier Design

115

Fig. 86 - Top GFRP Layer

A wooden board was built to make the slab surface perpendicular to

the tendons. This operation was performed in order to ease the post-

tensioning phase. An alternative solution would have been to build a

skewed slab, but it would have implied the development of a more

sophisticated tool to pull the tendons. The position of the wooden

triangles was designed in order to have the two tendons full centering

each board. After the post-tensioning, the slab portion coming out of

the abutments had to be cut (see Fig. 87 and Fig. 88).

116

45.7

38.445.7

38.4

43.7

52.1

43.7

52.1

7.9 m

6.1 m

50°

45.7

38.4

45.7

38.4

38.4

45.7

38.4

45.7

38.4

45.7

38.4

38.4

45.7

45.7

38.445.7

38.4

45.7

Fig. 87 - Board Detail

Fig. 88 - Wooden Board (the wire indicates the position of the tendons)

On August 23rd the plastic ducts were placed and tied to the GFRP

rebars as prescribed in the design specifications. Thirteen additional

117

safety ducts were placed (as shown in Fig. 89 and Fig. 90), straight

and carefully tied to avoid their floating on the liquid concrete during

the pouring.

22.86

15.24

0.038

0.177

1.22

0.254

bottom chairs

13 GFRP (G4)

13 GFRP (G3)

n.8 19 GFRP (G1)

n.8 19 GFRP (G2)

n.5 9 CFRP (C1)

up chairs (G5)

45.72

safety ductwork

plastic ductwork

0.039

F

F

F

F

F Fig. 89 - Slab Section

Fig. 90 - Plastic Ducts Detail

T connectors were used on each end (see Fig. 91) in order to allow the

injection of the grout as specified in the construction specifications in

Appendix D.

118

Fig. 91 - T Connectors Detail

Strain gages were also attached on the GFRP bars (see Fig. 92) and

they were positioned as shown in Fig. 93.

Fig. 92 - Attaching of a Strain Gage

119

Strain Gages Scheme

GFRP Rebars

5th from the right edge

the right edge 17th from

5.00

1.67

Upper Layer

Lower Layer

Midspan

1st wall axis

2nd wall axis

0.611.83

Fig. 93 - Strain Gages Scheme

The bridge deck was poured by the City Workers on August 25th. The

pour began at 7.30 am and was finished at 9.30 am (see Fig. 94).

Fig. 94 - Pouring Fragments

In order to let the concrete filling all the voids inside the FRP cage, a

more liquid concrete was used (Slump Test = 9 cm, ~4.5 in.), with the

addition of super-plasticizers which increase the slump without

120

detrimental effects on the concrete strength. A total of 16 concrete

cylinders were also prepared (see Fig. 95). Fig. 96 shows the just

poured slab.

Fig. 95 - Concrete Cylinders Execution and Slab Leveling

Fig. 96 - Poured FRP Bridge Deck

121

5.4.4 Post-tension of the Slab

After a week of curing of the slab, on September 1st the CFRP tendons

were post-tensioned. The pulling was achieved by means of the

machine already used for the two test specimens. Modifications were

required afterward in order to decrease its weight and improving its

functionality. These improvements were accomplished with the help

of Mr. Max Vath, a Rolla Technical Institute instructor. Moreover,

some handles were joined to better and easily transport it from a

tendon to another (see Fig. 97).

Fig. 97 - Pulling Machine Before and After the Optimizations

Fig. 98 shows the modified pulling device. It comprises an open steel

box having enough room to push the wedges inside the chuck after

pulling the tendon, an hydraulic jack to apply the pulling force, a

round steel plate, a load cell to measure the load, a second plate and a

second chuck. Fig. 99 details the terminal part of the pulling machine.

122

Fig. 98 - New Pulling Machine Ready for the Use

Fig. 99 - Load Cell and External Chuck Detail

Before pulling the tendons, they were carefully cleaned from grout,

grease or dust; then the wedges were pushed inside the external chuck

in order to provide a grip, using a copper sleeve (shown in the 3rd

section). The tendons were pulled by means of two hydraulic jacks,

connected to two pumps using load steps of 13-22 kN (3-5 kip) per

side. Also the use of a single pump with two jacks connected in series

was considered, but it was found to be harmful for the tendons being

OPEN STEEL BOX

HYDRAULIC JACK

INNER CHUCK

STEEL PLATES

LOAD CELL

OUTER CHUCK

LOAD CELL WEDGES

BARREL

TENDON

123

the applied load not uniform along the tendon. A special two-parts

steel hammer was built to push the wedges into the barrel (see Fig.

100).

The applied load was measured using a data acquisition system

(Orange Box) connected to a computer allowing to monitor the

applied load in real time (see Fig. 101).

Fig. 100 - Hammer Detail and Use on the Site

Fig. 101 - Orange Box

After reaching the desired load of 62.3 kN (14 kips), corresponding to

the 45% of the maximum allowable load to avoid the breaking of the

124

tendons for the tight grip, the wedges were pushed inside the inner

chuck, so the jacks were released, engaging in such way the inner

chucks. At this point the pulling device could be removed by cutting

the FRP bar with an electric saw (see Fig. 102).

Fig. 102 - Wedges Insertion and Cutting of the Tendon

In 2 weeks the first half of the tendons was pulled. Some delays

occurred due to further changes of the pulling machine in order to

better suit the slab surface: it was not always possible to have a

surface perfectly perpendicular to the tendon.

Fig. 103 - First Set of Pulled Tendons

125

The injection of grout followed the pulling of the tendons and was

carried out with the use of a pump, after sealing the chucks to avoid

grout leaking (see Fig. 104).

Fig. 104 - Grout Injection and Slots Locking

After 4 days of curing the inner chucks were removed drilling the

tendons inside of those.

On October 7th the second set of tendons was pulled after a stop

intermission of two weeks for a scheduled load test that the contractor

had to carry out in Winsconsin, then the works could restart; thus the

grout was injected also in the safety duct.

Eventually the extra part of the slab added for the post-tension was cut

using a big cutter, as shown in Fig. 105. On October 15th the

Southview bridge-deck was completed (see Fig. 106).

126

Fig. 105 - Cutting of the Slab Edge

Fig. 106 - Southview Bridge-Deck Completed

127

6 CONCLUSIONS

The present thesis deals with a new technology, explored with the

literature review, validated experimentally and verified on the field; it

is part of a series of collaboration activities between the University of

Naples, “Federico II”, and the University of Missouri-Rolla, UMR.

The project developed is a further step in the study and use of FRP in

civil engineering; in fact, the use of FRP both for mild reinforcement

(Glass FRP) and for post-tensioned reinforcement (Carbon FRP) was

required for the construction of a bridge deck in the City of Rolla,

Missouri.

The objectives of the project at UMR were as follows:

1. Evaluate the feasibility, behavior, and effectiveness of the new deck

system, showing how FRP, in the form of GFRP as passive and CFRP

bars as active internal reinforcement, could be an excellent solution

replacing the steel.

2. Provide analytical data in support of the enhanced shear capacity of

the concrete slab due to the CFRP prestressing.

Informations on existing prestressing and non-prestressing FRP bridge

decks have been given in the section two, so that they can be

compared with the new deck system that is the subject of this thesis. A

summary of the main research works on shear behavior of prestressed

FRP was also given, showing the lack of research projects on the

specific topic of this thesis.

Section three dealt with pre-construction investigations that were

conducted on two specimens representing a deck strip 457 mm (18 in)

wide and 7 m (23 ft) long, with the same geometry and amount of

128

reinforcement. They were built and tested one to investigate the

flexural behaviour, the other one the shear behaviour. The two beams

were constructed by the contractor peculiar for the project also

allowing for his familiarization with the use of non-conventional

materials. Their testing as continuous slabs over three supports

allowed the validation of the design calculations in terms of flexure

and shear capacities.

The specimens were reinforced using 3 f19 (6/8 in) GFPR bars as top

and bottom mat and 2 f9 (3/8 in) CFRP bars as prestressed tendons.

The position of the prestressed tendons was varied along the slab in

order to match the moment demand. In addition, in order to reproduce

the actual field conditions also f13 (4/8 in) GFRP bars spaced 305

mm (12 in) on center were placed in the transversal direction as

temperature and shrinkage reinforcement.

The experimental results were compared with the theoretical moment

and shear capacities of the slabs computed according to ACI 440.1R-

03 provisions and to the equation developed by Tureyen A. K. and

Frosh R. J.

According to ACI 440, the shear capacity increased from 30.7 kN (6.9

kip) with mild reinforcement to 69.8 kN (15.7 kip) adding the post-

tensioned reinforcement, while the flexural capacity increased from

76.0 kN-m (55 kip-ft) to 97.9 kN-m (72.2 kip-ft), showing that the

prestressing material is mostly needed for shear purpose rather than

for flexure.

On the basis of the experimental investigation, it can be concluded

that GFRP bars as passive and CFRP bars as active internal

reinforcement could represent a feasible solution replacing the steel

reinforcement of concrete slab bridges.

129

The objective of section four has been to provide the structural

analysis of the new FRP concrete bridge deck based on the AASHTO

HS20-44 design truck and provide calculations for its design using a

combination of non-traditional corrosion-resistant composites

materials.

Two load configurations were considered, consistent with AASHTO

Specifications, the HS20-44 design truck and the design lane; the

concentrated load and uniform load were considered to be applied

over a 3.0 m (10ft) width on a line normal to the center line of the

lane. These loads were placed in such positions within the design lane

as to produce the maximum stress in the member.

Since the design truck analysis produced the highest stresses on the

slab, only moment and shear related to this analysis were considered.

The design of the internal FRP reinforcement was carried out

according to the principles of ACI 440.1R-03, using the same

percentage of reinforcement already used ant tested in the

experimental phase.

Consistent with section 3 it was noted that the prestressing force is

most needed for shear purpose rather than flexure. In fact, if the

prestressing action was not considered, the concrete contribution to

the shear capacity of the slab would have been Vc=220.4 N/m (=15.1

kips/ft), and the final factored shear capacity of the bridge would have

been equal to φVn=86.09 kN/m << Vu=148.9 kN/m (φVn=5.9 kips/ft

<< Vu=10.2 kips/ft), that is the shear capacity of the slab with post-

tensioning; therefore the post-tensioning allowed to increase the shear

capacity of the slab more than 40%.

130

Section five detailed the installation of Southview bridge deck,

focusing on the post-tensioning of the slab, the most considerable and

crucial part of the project.

The slab is 25 cm (10 in) thick, 12 m long (40 ft) long and 6 m (19.83

ft) wide. It is supported by three intermediate reinforced concrete

vertical walls. The three spans lengths are, from North to South, 2.89

m (9.49 ft), 3.38 m (11.10 ft) and 3.30 m (10.82 ft), respectively, on

centers.

The erection of the substructure started the 21st of July; the extension

of the existing abutments and walls was performed by the City of

Rolla employees prior to the slab construction.

On August 17th the installation of the deck started, after the partial

embedding of anchoring GFRP bars into the central wall, in order to

provide a fix central section, avoiding the total slipping of the

superstructure during the prestressing operations. Hence, after the

laying of formwork, the neoprene pads on the two external walls and

on the abutments were placed in order to allow the axial sliding of the

middle fixed slab.

The installation of bottom and top longitudinal and transversal GFRP

reinforcement was completed by the 21st of August, after which the

GFRP reinforcement for the barrier was also placed on the ”FRP

side”.

Wooden boards were positioned in order to ease the post-tensioning

operations, then the placing of the ductwork, the T connectors and the

CFRP tendons inside the ductwork was carried out by August 23rd.

Before pouring, some strain gages were also attached on the GFRP

bars of the slab, in order to allow checking the behavior of the slab in

service.

131

The bridge deck was poured by the City Workers on August 25th; after

a week of curing, on September 1st the CFRP tendons were post-

tensioned.

The pulling and releasing of the tendons were the newest and the most

critical parts of the project.

The pulling of the tendons was achieved by means of the machines

already used for the two test specimens, with two hydraulic jacks and

special chucks to provide the grip between the machine and the

tendons themselves, as detailed in the third section.

On September 15th, after some delays occurred due to further changes

of the pulling machine in order to better suit the slab surface, the first

half of the tendons had been pulled, hence the grout was injected

inside their ductwork.

The same procedure was carried out for the second half of the

tendons; hence all the tendons edges outside of the slab were cut after

the curing of the grout.

On October 15th the edges of the slab were also cut.

The main result of this project has been to show how FRP, in the form

of GFRP as passive and CFRP bars as active internal reinforcement,

could be a feasible solution replacing the steel reinforcement of

concrete slab bridges, and specifically the enhanced shear capacity of

the slab due to the CFRP prestressing.

Moreover, the following conclusions can be summarized:

• The real advantage in the use of FRP materials as reinforcement

of a bridge deck is seen through its increased durability, mainly

due to the absence of corrosion. To better show it, a barrier with

GFRP reinforcement was built on the new “FRP side”, in order

to have a comparison during time with the steel reinforced

132

concrete barrier on the opposite side, which was going to be

built after the FRP bridge deck installation.

• Utilizing FRP in the form of reinforcing bars allows for the use

of many steel-RC concrete practices. The fabrication and

installation details were nearly identical to the methods

regularly utilized for steel-reinforced slabs.

• The installation of the bridge highlighted the fact that having an

efficient system is as important as having the adequate

components. As well, for a new technology its learning curve

must be overcome before its applications can be conducted

proficiently.

• There was an increase in productivity using the FRP materials

in each stage of construction versus conventional steel. This is

directly associated with a labor savings. While this labor

savings does not offset the initial increase in material costs, it is

hoped that will become a factor in lowering the total life cycle

cost of the bridge.

• Further investigation must be done in this field in order to

improve and refine the techniques related to the post-tensioning

of bridge-decks using CRFP, its future development being

certainly hopeful and helpful.

design and construction of a bridge deck using frp as … · a research project was undertaken to...

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