1resistance of steel connections to low-cycle fatigue

11th European Conference on Earthquake Engineering 1998 Balkema, Rotterdam, ISBN 90 5410 982 3

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Resistance of steel connections to low-cycle fatigue

A. Plumier & M.R.AgatinoDepartment of Civil Engineering, University of Liege, Belgium

A. Castellani, C.A. Castiglioni & C. ChesiDepartment of Structural Engineering, Politecnico di Milano, Italy

Keywords: Earthquake, Steel, Welding, Frame, Connection, Ductility, Fatigue, Experimental Tests

ABSTRACT: following the northridge and Kobe earthquake, attention of researchers has been fo-cused on the problem of the brittle collapse of beam to column connections in moment resistingsteel frames. large evidence has been given by the earthquake to this phenomenon which, althoughnot involving the frame global collapse, drastically changes the expectations about energy dissipa-tion and ductile behaviour of connections. the paper presents an experimental testing programwhich has been planned with the purpose of contributing to a deeper knowledge of the followingproblems: the low cycle fatigue behaviour of welded joints and the implications of brittle failure inwelded connections on the global frame response.

1 INTRODUCTION

In the variety of cases of structural damage which were observed after the Northridge earthquake, amost prominent one occurred in steel moment resisting frames (Bertero et al. 1994, Tremblay1995); although not reaching collapse, this kind of frames presented brittle failures at moment con-nections in welded beam to column joints. The obvious relevance of this kind of behaviour has ledto a detailed mapping of such cases. Although similar failure situations had already been observedin previous earthquakes and during laboratory experiments, the Northridge earthquake has givenspecial evidence to the problem of brittle failure of welded joints and to the need of clarifying theconditions under which this kind of failure is likely to be induced by seismic actions. This impliesthe study of the low cycle fatigue behaviour of beam to column connections. The same problem isbeing addressed as it is of particular interest, at present, in the European Community Countries aswell, where common codes have recently been developed covering the various design fields,among which the design of steel structures (CEN, Eurocode 3, 1994) and the design of buildings inseismic areas (CEN, Eurocode 8, 1994). Codes are now subject to public debate and contributionsshould be provided on critical design problems, like the case of welded joints subject to seismicactions. In this context, funding has been accorded by the European Community for a wide project,named Steelquake, which is under way at the moment and which is presented in this paper. TheSteelquake project concerns the analysis of the behaviour of civil engineering steel structures, ofthe moment resisting category, under earthquake loads. The overall objective is to provide a betterinsight into the actual behaviour of civil engineering steel structures, of the moment resisting cate-gory, under earthquake loads, implementing in such an analysis a quantitative reference to the fail-ure of connections in terms of low cycle fatigue, considering the duration of the earthquake and thecorresponding number of cycles supported in dissipative (plastic) zones. The structural behaviourfactor characterising the structure (the so-called "q" factor) presently proposed in Eurocode 8, is notrelated to the fatigue resistance of the potential failure zones, that is beam to column connections.The aim is to allow the full design (weld size, thickness of components, dimensions) of steel frameconnections, taking into account the duration of the earthquake and the corresponding number ofcycles to be supported in dissipative (plastic) zones, that is to introduce q factors depending on theconsidered earthquake time-history. A classification of connections, in terms of q values, is ex-


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pected as a result of the project research. In order to achieve the above-said objective, full scaletesting of beam to column connections and testing on welds are carried out on different specimensin Milan and Liege; shaking table tests at Athens and at ISMES Bergamo and pseudo-dynamic testsat JRC Ispra are performed to provide complementary experimental data concerning the global be-haviour of steel frames; parallel numerical analyses are performed with an aim to deeply interpretthe experimental behaviour both of the investigated single connections and steel frames.

The project has been conceived with the purpose of providing some contributions to two exist-ing problems: 1) a clear understanding of the real behaviour of welded beam to column connec-tions, a problem which still deserves accurate investigation despite of the work done in this area byseveral researchers (Plumier 1994, Bernuzzi et al. 1997) and 2) the interpretation of the influenceof the connection failure on the global frame answer to seismic actions; in the Northridge earth-quake no remarkable effect was noticed on the global frame behaviour as a consequence of jointcollapse. In this paper, only the research activities related to phase 1 is presented.

2 BEAM TO COLUMN CONNECTIONS AND WELDING PROCEDURES

2.1 Specimens selection and test set-up

The first part of the project, discussed here, is focused on the experimental response of weldedconnections, for which the low cycle fatigue problem is investigated. The specimen selection aimedat covering a significant variety of beam to column connections, frequently adopted in both Euro-pean and American practice. Four typologies, as shown in Figure 1, were identified for this pur-pose, based on welded joints (type C) or partially welded joints (type A). All of these provide inter-esting cases; note, in particular, connection type A2, based on the concept of shifting the plastichinge zone away from the node, a joint design approach which has been recently proposed in theU.S., and for which some results are already available, based on work of Ballio & Castiglioni(1994).

Figure 1. Beam to column connection typologies without (left) and with (right) node reinforcing plate.


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Figure 2. Different welding procedures C1 and C3: with steel backing bar; C2: with polyester backing bar.

Figure 3. Beam to column connection Figure 4. Full scale-small size specimensfull scale specimen. for weld size testing.

The presence of stiffening plates in the panel zone of the column is also to be investigated; forthis purpose, each of the above typologies include one additional specimen, with a node stiffeningplate. The research, however, is mainly focused on the beam-column connection behaviour; for thisreason, an external T-shaped joint has been selected for the experimental program, in which col-lapse is expected to occur preferably at the beam-column interface. For connection type C differentwelding procedures have been considered. These include the full penetration double bevel grooveweld and the full penetration single bevel groove weld with a steel or polyester backing bar, Figure2. A V-shaped steel backing bar is also considered (C3). Some special care has been devoted towelding execution, which was done by competent structural steel fabricators, using certified weld-ers.

A typical view of the full-scale specimens is shown in Figure 3. During tests, the column is kepthorizontal, hinged at both ends, and the (horizontal) loading system applied at the top of the beam.A total of 32 specimens have been fabricated, 16 specimens (type C1, C2 and C4) to be tested inMilan and 16 specimens of types A, B and C3 to be tested in Liege. Testing has been completed. Inaddition to the beam-column joints, full-scale small size specimens have also been fabricated withthe purpose of verifying whether the weld preparation and the specimen sizes are critical parame-ters. The specimens (Fig. 4) incorporate welded connections representing the welded beam to col-umn flange connection. A total of 40 specimens have been constructed and subjected to cyclictesting.


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2.2 Testing procedure

In 1986, a recommended testing procedure was proposed by ECCS for cyclic testing of steel mem-bers and joints. The ECCS recommendations (ECN, 1986) proposed to perform cyclic tests ap-plying to the specimen cycles of increasing amplitude with at least three cycles for the same am-plitude, similarly to what is proposed by ATC 24 (1992).

Since then, a number of research programs were carried out in various European countries,adopting such ECCS recommended procedures. Tests were carried out in Italy, Germany, Belgium,Portugal and France, on members, connections and structural subassemblages. In particular a re-search project was carried out with the sponsorship of Arbed Research on beam to column connec-tions, and tests were performed at different research centers in various European countries; see, forinstance, Ballio & Chen (1993). After nearly ten years, some criticism was raised with regard to the1986 ECCS recommendations. In particular, with regard to the Testing Procedures, it has been rec-ognised that:

1. Repeating three cycles for each cycle amplitude and then increasing the amplitude does notprovide direct information regarding the damage accumulation and strength degradation corre-sponding to one particular ductility demand; in fact, at increasing amplitudes corresponds an iso-tropic strain hardening effect resulting in an increment of the load carrying capacity of the member,but this effect is opposite to the strength degradation due to local buckling and low cycle fatigue,which are directly connected to the cycle amplitude and the number of imposed cycles;

2. Sometimes tests were performed under unusually large cycle amplitudes, i.e. simulatingsituations and ductility demands unlikely to occur during earthquakes.

It was then proposed by various authors both in Europe (Ballio & Castiglioni 1994, Bernuzzi etal. 1997) and in the U.S. (Krawinkler & Zohrei 1983) to perform tests with cycles of a constantamplitude; such procedures present, with regard to the ECCS recommended testing procedure, theadvantage of allowing a clear understanding of the damage accumulation process as well as of theparameters governing it.

The scope of the research is the development of a cumulative damage model for assessing theperformance of structural components under arbitrary loading histories and evaluating the effects ofinelastic cycles on a limit state of acceptable behaviour. A cumulative damage model is generallybased on a damage hypothesis and includes structural performance parameters to be determinedexperimentally (Bernuzzi, et al 1997, Krawinkler & Zohrei 1983). Hence a multi-specimen testingprogram is required, encompassing constant amplitude loading tests on identical specimens. Foreach test, a new specimen must be used, since each specimen is to be tested to failure under defor-mation amplitudes covering the range of performance assessment.

Cyclic tests are performed following a constant amplitude loading history in the plastic range,after a few cycles in the elastic range. In particular, it is proposed to adopt the loading historyschematically represented in Figure 5, vy being the normal yield displacement, to be computedtheoretically by means of usual structural mechanics.

Figure 5. Schematic of the loading history.


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Two cycles will be performed for each cycle with a displacement range v < 2 vy, while cyclesin the plastic range will be continued until complete failure of the specimen. Various values of themaximum displacement range are considered: v = 4 vy , 6 vy , 8 vy. Experience gathered in pre-vious testing programs, when it was assumed a cumulative damage model based on an S-N curveapproach, indicated that such procedure allows for a correct identification of low cycle fatiguecurves for each typology. To run the test, a definition of the failure of the specimen must be givenin cyclic testing, in particular when cycles of constant displacement amplitude are applied, thefailure may be very long process if a zero resistance is chosen as the definition of failure. Refer-ing to practical applications in structures, it is clear that the structure is out of use before a zero re-sistance is reached. Conventionally in our testing, failure is determined by a drop in resistance of50 % of the initial value obtained at the same displacement.

3 PRESENTATION OF LOW-CYCLE FATIGUE TEST RESULTS

The parameter characterizing high cycle fatigue is the number N of cycles at failure and the stresstange , defined as follow for overall elastic deformations:

W

dP

P

P

W

dP

W

M === (1)

v

v

P

P = (2)

W

dP

v

v = (3)

A working hypothesis is developed for low cycle fatigue. is the basic parameter characterizingfatigue; in the dissipative zone, an equivalent stress * can be computed as:

W

dP

v

vE y

y

== * (4)

As test on each specimen are run at constant v and as Py and vy are known, it is easy to compute* and to present the results in typical diagrams used for the presentation of high cycle fatiguewith * in ordinate and the number N of cycles to failure as abciss.

Expressed in the way, the results allow the definition of low cycles fatigue (Whler) curveswhich can be compared to curves concerning the resistance to high cycle fatigue (classical). In thepresentation of the test results, reference is made to the design curves of Eurocode 3 (design ofsteel structures).

Figure 6.


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4 TEST RESULTS

All the tests in Lige and in Milan were constructed in early 1998. Some first significant results arepresented hereafter, but the analysis of the results is still going on, involving numerical evaluationand considering experimental work still going on at the other testing sites.

4.1 Specimens type A1

Specimens A1 correspond to a rigid full strength bolted connection. The end plate of the beam iswelded with K preparation. The failure takes place in the beam: flange buckling or cracks in thebeam close to the weld. The results are presented in terms of low cycle fatigue (*, N) (Fig. 7).Results A15 (95 mm) is situated on a higher curve than A12 (38 mm) and than A13 (57 mm). Thiscorresponds to two different failure modes: without apparent deformation in the flange for A12 andA13, with buckles in the flange for A15. The result of specimen A1B2 (33 mm), which had nodouble plate in the web of the column, lies higher than A12 (38 mm); this may be explained by theexistence of yielding of the panel zone that brings an increase in resistance to low cycle fatigue.Globally, it can be concluded that connection A1, which is typical European design involvingwelding at the shop and bolting on the site, has a very good behaviour. It corresponds to yielding inthe flange of the beam or to yielding in the panel zone of the column (specimen A1B).

4.2 Specimens type A2

The specimens A2 were subjected to several problems, which are not related to failure in welds: Failure of bolts in the original design with 4 bolts in the connection (specimens A2 I , A2 II) Lateral buckling, caused by a combination of web and flange buckling in the section where the

hunch starts (specimen A23); consequently the lateral support were increased in the testing rig Even with increased lateral support (specimen A25, A26) the combination of local buckling of

web and flange in the section where the haunch starts remains the factor governing failure; thiscan be explained by particular state of stress at this place; indeed the haunch introduces a thrustonto the web; this was known and complete web stiffener would be needed to really preventthe web buckling.

As a general comment on haunched beams the following can be expressed: Putting a haunch increases the rigidity and the resistance of the beam However, the haunch beam cost is very high because it requires much preparation of stiffeners,

a very high and thick plate and very long welds; the cost of connection A2 is about 80% higherthan the cost of A1.

The results are presented in terms of low-cycle fatigue in Figure 8.

4.3 Specimen B

Connection B is a semi rigid partial strength connection in which bolts are perpendicular to thebended beam. The bolts are prestressed to a 9 kNm torque.

It is intended to develop energy dissipation through the ovalization of boltholes and throughfriction between the web of the UPN 300 beam and the flange of the HE 300 M column. Bothmechanisms work:- friction gives a constant resistance throughout the displacement,- bearing resistance provides increases in resistance at both left and right side of the M - dia-

grams (Fig. 9).These mechanisms are not subjected to degradation of strength up to number of cycles which canbe considered very high (over 40) in the earthquake context.


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Figure 7. Whler curve for specimen type A1. Figure 8. Whler curve for specimen type A2.

Figure 9. M - diagram for specimen B2.

The imposed displacements in test of specimens B have been chosen on the basis of a maximum5 % drift of the building in which they would be present. In the range of displacements considered,the friction force is quite constant and independent of the maximum displacement. It is around 50kNm / (4 x 0,113 x 2) = 55 kN / bolt in both directions.

This last value is coherent for a M27 10.9 bolt with partial prestress and a friction coefficientaround 0,25. Once the bearing resistance is involved, the total resistance becomes higher, by a fac-tor of 4 at the beginning of the cycles; this correspond to a total resistance of about 4 x 55 = 220 kN/ bolt. This is very similar to the design resistance computed in the design of the specimens (217kN / bolt). In reality, the phenomenon explaining the resistance of the connection is more compli-cated than a simple addition of "friction" and "bearing resistance", because at the 1st cycle, theyielding realized by the ovalization of holes generates a local increase in the thickness of the webof the UPN 300, which brings two additional factors: an increased prestress because the bolt isforced to elongate and an increased friction because of the uneven surface produced by this "forg-ing" phenomenon. The complete understanding of the resistance mechanism in such a connectionrequires an extensive study and numerous small tests.


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4.4 Specimen C1

Specimen type C1 were realized with a full penetration single bevel groove weld with a steelbacking bar. Figure 10 shows the hysteresis loops for test C1-50, while Figure 11 shows the hys-teresis loops for C1-100 one. These specimens were tested under displacement ranges v = 50mm and v = 100 mm. The yield displacement vy being approximately 23-25 mm, these dis-placement ranges resulted in ductility ranges v/ vy = 2 and 4 respectively. These specimensare similar to those adopted for the moment resisting frame structures in which failures were re-ported during the Northridge Earthquake. Specimen C1-50 collapsed by failure of the beam flangeat the weld toe; this failure can be considered a brittle failure, because it took place suddenly, with-out noticeable deterioration in the load carrying capacity of the specimen. Only limited local buck-ling was evident at the end of the test in the beam flanges. Completely different was the behaviourof specimen C1-100, which collapsed by failure of the beam flange due to large plastic deforma-tions in the buckles, at the plastic hinge location. In the case of this specimen, the deterioration isobvious and represents a clear warning sign of an incipient collapse. The different behaviour of thetwo specimens is also evident at Figure 14, which presents the trend of the absorbed energy at eachcycle (E) normalized on the energy at the first cycle in the plastic range (E0). It can be noticed thatin the case of specimen C1-100, the ratio E/E0 shows an evident reduction during all the test dura-tion, and before the last cycles leading to failure, it is already reduced below a value of 0.50. On theother hand, the same ratio remains higher than 0.8 for specimen C1-50, dropping then suddenly, inthe last two cycles, below 0.2.

Figure 10. Hysteresis loops for specimen C1-50. Figure 11. Hysteresis loops for specimen C1-100.

In addition, it should also be noticed that the conventional failure for specimen C1-50 can beconsidered attained after 17 cycles in the plastic range, while that of specimen C1-100 was at-tained at cycle n. 12. This difference in the life of the specimens, that is very small, can be ex-plained only by the different failure modes. From these results, it seems that cycles with a smallamplitude can lead to brittle fracture and may be more dangerous because they provide much lessvisual warning about the degradation of the connection than cycles of large amplitude. These ones,on the contrary, induce the formation of a plastic hinge with local buckling giving a very ductilebehaviour and a large energy dissipation.

4.5 Specimens C2

Specimens C2 are identical to specimen C1 but for the material of the backing bars, which is poly-ester. All the observation made for C1 are relevant. The distinction between two behaviours, onebrittle (C2B50, C2B75) and one ductile ( C2B100, C2B125) is even clearer (Figs. 13, 15).


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Figure 12. Whler curve for specimen type C1. Figure 13. Whler curve for specimen C2.

Figure 14. Energy absorption in Hysteresis loops Figure 15. Energy absorption in Hysteresis loopsfor specimens type C1-B. for specimens type C2.

Figure 16. Whler curve for specimens C4.


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4.6 Specimens C3

Specimens C3 are similar to C1, but for the shape of the backing bars which are triangular. Thebacking bar tested in specimens C3 did not allow a complete butt weld to be made. Yielding wasrestricted to the weld material and it did not last long: 2 or 3 cycles, to be compared for instance to30 and 77 in specimens A1.

The high yield strength of the IPE 450A material (405 N / mm instead of the expected 300N/mm) may also be participating to this early failure, since the weld material had not been chosenspecifically for a yield strength higher than expected.

4.7 Specimens C4

Specimens C4 correspond to a K preparation and welding from both sides of the flanges, which areconditions for high quality welds.

Here again two different behaviours can be observed, to which two levels of Whler curves cor-respond, Figure 16:- Brittle for displacement up to 50 mm- Ductile above 50 mm

4.8 Comparison between various design

In the above table, the index of EC3 curves that would represent the behaviour of tested specimensis given. For each specimen type two values are presented, which correspond to the two differentmechanical behaviour observed: Low cycle fatigue excursions, no buckling, rather brittle failure High yield excursion, buckling, progressive decrease of resistance.

Table 1. Comparison for of EC3 fatigue line (MPa) between various designs.

SPECIMEN TYPE Small yield High yield displacement v = 2 vy displacement v = 4 vy

A1 42 50A2 / 40C1 21 56C2 32 63C3 10 18C4 29 63

5 CONCLUSIONS

The tests realized with the objective to define low cycle fatigue curves corresponding to variousdesign detail of connection used in moment frames allow several practical conclusions. The results obtained set forward the validity of a quantification of cyclic tests in term of low

cycle fatigue On that basis, the various connection design tested can be classified in terms of average of re-

sistance, but also of scatter; in this formulation, the word design includes the many factors in-fluencing the behaviour, like base material, weld material, groove type, welding sequence, typeof backing bar, flexibility of eventual end plates, yielding resistance of the panel zone of thecolumn

Two basic different behaviours have been observed, independently of the type of connectiondesign. One takes place when the plastic strains realized during the cyclic test are small,


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around two times the yield stains, and no buckling of the beam is observed; then, the strengthdegradation is small until a relatively sudden failure takes place. The other type of behaviourhappens when large plastic cyclic excursions are realized; then the resistance progressivelydrops, as buckling develops.

ACKNOWLEDGEMENTS

Financial support to the project has been provided by the European Community Commission forEnvironment and Climate Programme, Directorate XII, Science, Research and Development;Contract N. ENV4-CT96-0278; Proposal N. PL950671.

REFERENCES

ATC-24 1992. Guidelines for Cyclic Seismic Testing of Components of Steel Structures, Applied Technol-ogy Council.

Ballio, G., Chen, Y. 1993. An Experimental Research on Beam Column Joints: Exterior Connections,Giornate C.T.A., Viareggio.

Ballio, G., Castiglioni, C.A. 1994. Seismic Behaviour of Steel Sections, Journal of Constructional SteelResearch, Vol. 29, pp. 21-54.

Bernuzzi, C., Calado, L., Castiglioni, C.A. 1997. Behaviour of Steel Beam-to-Column Joints under CyclicReversal Loading: an Experimental Study, SSDS, Proceedings, Nagoya, Japan.

Bernuzzi, C., Calado, L., Castiglioni, C.A. 1997. Low Cycle Fatigue of Structural Steel Components: Meth-ods for Re-Analysis of Test Data and a Design Approach Based on Ductility, submitted for publicationon the Journal of Earthquake Engineering.

Bertero, V., Anderson, J.C., Krawinkler, H. 1994. Performance of Steel Building Structures during theNorthridge Earthquake, EERC, University of California, Berkeley, Rep. n. UCB/EERC-94/09.

European Committee for Standardization (CEN) 1994. ENV 1993-1-1 Eurocode 3: Design of Steel Struc-tures.

European Committee for Standardization (CEN) 1994. ENV 1998 Eurocode 8: Design Provisions for Earth-quake Resistance of Structures.

European Convention for Constructional Steelworks 1996. Recommended Testing Procedures for Assessingthe Behaviour of Structural Elements under Cyclic Loads, Technical Committee 1, TWG 1.3 - SeismicDesign, Publ. N. 45.

Krawinkler, H., Zohrei, M. 1983 Recommendations for Experimental Studies on the Seismic Behaviour ofSteel Components and Materials, The John Blume Earthquake Engineering Research Center, StanfordUniversity, Rep. n. 61.

Plumier, A. 1994. Behaviour of Connections, Journal of Constructional Steel Research, Vol. 29, N. 2, pp.95-119.

Tremblay, R., Timler, P., Bruneau, M., Filiatrault, A. 1995. Performance of Steel Structures during theNorthridge Earthquake, Canadian Journal of Civil Engineering, Vol. 22, N. 2, April.

1resistance of steel connections to low-cycle fatigue

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