JOINT DETAILING BETWEEN DOUBLE TEE BRIDGE GIRDERS FOR IMPROVED 1 SERVICEABILITY AND STRENGTH 2 3 4 5 Nadim Wehbe, Corresponding Author 6 South Dakota State University 7 Crothers Engineering Hall 120/Box 2219, Brookings, SD 57007 8 Tel: 605-688-4291; Fax: 605-688-6476; Email: nadim.wehbe@sdstate.edu 9 10 Michael Konrad 11 Kiewit Engineering Co. 12 3555 Farnam Street, Omaha, NE 68131 13 Tel: 402-943-1409; Email: Michael.Konrad@Kiewit.com 14 15 Aaron Breyfogle 16 South Dakota Department of Transportation 17 700 E. Broadway, Pierre, SD 57501 18 Tel: 605-773-3871; Fax: 605-773-4713; Email: aaron.breyfogle@state.sd.us 19 20 21 Word count: 4,411 words text + 12 tables x 250 words (each) = 7,411 words 22 23 24 25 26 27 28 Original Submission Date: July 30, 2015 29 Revised Submission Dtae: November 15, 2015 30

Wehbe, Konrad, Breyfogle 2 1 ABSTRACT 2 Precast/prestressed double tee bridge girders are widely used for accelerated bridge construction 3 on local roads in South Dakota. A common issue among existing double tee bridges in South 4 Dakota is the rapid deterioration of longitudinal joints (shear keyways) between adjacent girders. 5 Deteriorated joints allow moisture and de-icing chemicals to reach the deck reinforcement, leading 6 to premature corrsosion of reinforcing steel and spalling of concrete. The structural performance 7 of conventional and proposed longitudinal joints between precast double tee bridge girders was 8 examined experimentally under cyclic and monotonic loading. The proposed contiuous joint with 9 overlapping steel mesh reinforcement in a grouted shear keyway exhibited substantially improved 10 serviceability and strength performance characteristics over the conventional grouted joint with 11 discrete welded connections. 12 13 14 Keywords: ABC, Bridge Deck, Double Tee, Shear Key, Fatigue, Strength 15 16

Wehbe, Konrad, Breyfogle 3 INTRODUCTION 1 Precast bridge superstructure elements are essential for accelerated bridge construction. Due to 2 their ease of construction and reduced construction time and cost, precast/prestessed double tee 3 bridge girders are routinely used by local governments in South Dakota for rapid construction of 4 bridges on local roads. Detailing of longitudinal joints between precast bridge girders for adequate 5 shear transfer remains a major concern especially in “decked” precast girders, such as double tee 6 girders, which do not require cast-in-place bridge decks. 7

The conventional joint detailing used for double tee girder bridges in South Dakota 8 consists of discrete welded connections spaced along a grouted longitudinal joint (shear keyway) 9 between adjacent girders. A common issue among existing double tee bridges is that the 10 longitudinal joints deteriorate with time, most likely due to inadequate shear connection between 11 adjacent girders. It is only a matter of time before the grout begins to crack along the joint, 12 creating a path for moisture and de-icing chemicals to reach the steel reinforcement in the deck, 13 and leading to corrosion, concrete spalling, and structural degradation of the bridge. Short-term 14 maintenance such as asphalt overlays can temporarily seal longitudinal joints, but asphalt overlays 15 are costly and have tendency to form reflective cracks directly above the longitudinal joints. 16 Figure 1 shows reflective cracking in the asphalt overlay of a double tee bridge deck and 17 deterioration of a double tee flange at the longitudinal joint. 18 19

FIGURE 1 Deterioration of double tee bridge girders at longitudinal joints. 20 21 PREVIOUS WORK 22 Jones (1) investigated two types of longitudinal joint connection for double tee bridge girders used 23 by Texas Department of Transportation (TxDOT). One type was a standard TxDOT detail and the 24 other was a proposed simple detail for future use. Both types incorporated discrete welded 25 connections and a grouted shear keyway. Laboratory tests of full-scale specimens revealed 26 adequate performances of the standard joint under static loads and the simple proposed joint under 27 both cyclic fatigue and static loading. The study did not investigate water tightness of the joints. 28

Li et al. (2, 3, 4) conducted a series of studies involving development of improved 29 longitudinal joint details in decked bulb tee girders. The proposed joint details relied mainly on 30 non-shrink grout and steel reinforcement to provide the joint with flexural strength under 31 transverse bending. The proposed details were tested using small slab specimens, rather than 32 girder specimens, joined with the proposed longitudinal joints. The slab specimens were subjected 33 to cyclic fatigue and static loading. The test setup may have not completely simulated the true load 34 and deformation demands at these joints under realistic boundary conditions. 35 36

Reflective Cracks

Corrosion

Concrete Spalling

Wehbe, Konrad, Breyfogle 4 RESEARCH SIGNIFICANCE 1 Due to rapid deterioration and increased traffic demands, many bridges on the South Dakota local 2 highway system need replacement. The desired rate of bridge replacement created a backlog of 3 local bridges in need of replacement. Double tee bridge girders provide economic and rapid 4 construction technique for bridge replacement. Although the service life of double tee girders used 5 on local roads was expected to last 50 to 70 years, some double tee bridges built less than 40 years 6 ago already need replacement due to premature deterioration caused by inadequate longitudinal 7 joints. This experimental study was performed in order to develop and verify the performance of a 8 simple joint detailing for enhanced serviceability and strength. 9 10 EXPERIMENTAL WORK 11 Two 40 ft (12.2 m) long full-scale bridge superstructure specimens, each consisting of two joined 12 double tee girders, were tested at the Lohr Structures Laboratory at South Dakota State University 13 (SDSU). Each specimen represented two adjacent interior girders of a 33 ft (10.06 m) wide bridge. 14 One specimen, labeled “Conventional”, incorporated the longitudinal joint detailing that has been 15 traditionally used in South Dakota (grouted keyway with discrete welded steel connections). The 16 other specimen, labeled “Proposed”, incorporated a redesigned continuous longitudinal joint with 17 a grouted shear keyway that is 4 in. (102 mm) wider than the conventional shear keyway. The 18 redesigned joint does not require welded connections, but is reinforced with overlapping wire 19 mesh that is extended out of the decks. The “Proposed” specimen was tested with and without 20 precast concrete diaphragm placed between stems of adjacent girders to restrain transverse rotation 21 of the joint. Details of the specimens are presented hereafter. The main objectives for the 22 laboratory tests were to evaluate serviceability and strength of the conventional and the proposed 23 longitudinal joints when subjected to both fatigue (cyclic) and increasing monotonic loading. 24 25 Test Specimens 26 Figure 2 shows a cross section of the standard 46 in. (1168 mm) wide by 23 in. (584 mm) deep 27 double tee prestressed units used in this study. The flange was 5 in. (127 mm) thick and reinforced 28 with two layers of Grade 70 (483 MPa) 4 x 8 - D8 x D4 (ϕ 8.1 mm @ 102 mm c/c x ϕ 5.7 mm @ 29 203 mm c/c) welded wire mesh and four #4 (ϕ 12.7 mm) bars. The stem tapered from a width of 30 6.25 in. (159 mm) at the top to 5 in. (127 mm) at the bottom. The stem reinforcement consisted of 31 six 0.5 in. diameter (12.7 mm) 270 ksi (1863 MPa) low lax 7-wire strands, each tensioned to 31 32 kips (138.3 kN). The shear reinforcement in the stem consisted of one layer of 4 x 8 D8 x D4 33 welded wire mesh. This girder section represents the predominant section size used for 34 replacement of low volume bridges in South Dakota. 35 36

37 FIGURE 2 Cross section of standard 23 in. (584 mm) double tee girder. 38

Wehbe, Konrad, Breyfogle 5

The girders were cast at a local precasting facility in South Dakota after being instrumented 1 by SDSU personnel. The measured concrete strengths on the day of testing for flexural strength 2 were 9,140 psi (63.1 MPa) and 8,570 psi (59.1 MPa) for the “Conventional” and the “Proposed” 3 specimens, respectively. Using AASHTO equations (6), the corresponding theoretical flexural 4 strengths for single girders were determined as 584 kip-ft (794 kN-m) and 580 kip-ft (789 kN-m) 5 for the “Conventional” and the “Proposed” specimens, respectively. The test specimens were 6 assembled in the laboratory following the procedure used in field construction. The 7 “Conventional” and the “Proposed” specimens were identical except for the longitudinal joint 8 detailing. 9 10 “Conventional” Specimen 11 Details of the “Conventional” specimen are shown in Figure 3. The longitudinal joint consisted of 12 a grouted keyway that extended the length of the girder and incorporated welded steel connections 13 at 5 ft (1.52 m) center-to-center spacing along the girder length. The welded steel connection 14 consisted of a ¼ in. (6.4 mm) thick steel plate field-welded to a pair of steel angles, one in each 15 girder, which were anchored in the flange concrete during fabrication. The steel angle was 1.5 in. 16 x 1.5 in. (38 mm x 38 mm) by 6 in. (152 mm) long, with two 5 in. (127 mm) long headed studs 17 welded to the angle for anchorage into the concrete. Welding of the plate to the angles was 18 performed at the laboratory by a certified welder using 3/8 in. (9.5 mm) fillet weld. After all the 19 steel connections were welded, the keyway was grouted using non-shrink grout with a minimum 20 28-day compressive strength of 4500 psi (31.1 MPa) as specified by South Dakota Department of 21 Transportation (5). The grout used was non-metallic, non-corrosive cementitious grout with 22 controlled net positive expansion. The measured 28-day compressive strength varied between 23 5,630 psi (38.8 MPa) and 6,650 psi (45.9 MPa) for the grout batches used at different stages of the 24 experimental testing. To allow for visual inspection of the connections during the test, 6 in. (152 25 mm) long gaps at the welded connections were left un-grouted. 26 27

Plan View

Section B-B

Section A-A

Section C-C

FIGURE 3 “Conventional” specimen details. 28 29

Wehbe, Konrad, Breyfogle 6 “Proposed” Specimen 1 Figure 4 shows the “Proposed” specimen’s joint details. The longitudinal joint was 4 in. (102 mm) 2 wider than that of the “Conventional” specimen and was reinforced with overlapping 4 x 8 - D8.0 3 x D4.0 welded wire mesh that extended from the deck reinforcement for a distance of 6 in. (152 4 mm). The wire mesh extension was provided during the fabrication of the girders. Thus, the joint 5 reinforcement consisted of 0.319 in. (8.1 mm) diameter deformed wires spaced at 4 in. (102 mm) 6 center-to-center for a total steel area of 0.24 in2/ft (1666 mm2/m). Two 0.25 in. (6.4 mm) diameter 7 longitudinal bars (lacer bars) were added to the overlapping mesh in order to develop the joint 8 reinforcement. According to AASHTO (6), two longitudinal wires spaced at least 2 in. (51 mm) 9 apart are sufficient to develop the deformed wire in the transverse direction. The longitudinal joint 10 was then grouted the entire length of the girder using non-shrink grout with a specified minimum 11 compressive strength of 4,500 psi (31.1 MPa). The construction of the longitudinal joint required 12 temporary plywood to be placed at the bottom of the longitudinal joint while grout was being 13 placed. The “Proposed” specimen was tested with and without an option to restrain the rotation of 14 the girders relative to one another. The restraint was accomplished by means of a diaphragm 15 assembly placed between interior stems of adjacent girders. The assembly consisted of a 6 in. x 12 16 in. (152 mm x 305 mm) concrete cylinder strut to restrain the closing of the gap at the bottom of the 17 stem, and a ¾ in. (19.1 mm) diameter tie bolt to restrain widening of the gap between the stems. 18 The cylinder ends were chamfered to allow for a snug fit between the stems. A 1 in. (25.4 mm) 19 diameter polyvinyl chloride (PVC) sleeve at the center of the concrete cylinder allowed for 20 placement of the tie bolt through the cylinder. Galvanized steel sleeves in the stems allowed for 21 the passage of the steel bolts. The diaphragm assemblies were placed at 5 ft (1.52 m) on center 22 along the length of the girder. 23 24

25 FIGURE 4 “Proposed” specimen joint details. 26 27 Test Setup and Procedure 28 The test setup, shown in Figure 5, was identical for both test specimens. Each specimen was 29 simply supported on concrete reaction blocks. The effective span between the supports was 39.17 30 ft (11.94 m). Elastomeric bearing pads were placed under the girder stems. The girders were 31 labeled “A” and “B”, with girder “A” being the loaded girder and girder “B” being the trailing 32 girder. Loading was applied at the mid-span of girder “A” by means of a 328 kip hydraulic 33 actuator. The load was spread over a 10 in. x 20 in. (254 mm x 508 mm) loading steel plate that 34 was placed at the edge of the longitudinal joint in order to maximize the shear stresses through the 35 shear keyway. 36 37

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1 FIGURE 5 Test setup. 2 3

A load cell was placed under each of the four stems at one end of the specimen to measure 4 load distribution to the supports. Linear variable differential transformers (LVDTs) were used to 5 measure the mid-span deflections of girders A and B, the vertical movement of girder A relative to 6 girder B across the longitudinal joint, the transverse rotation of the joint, and the deflection 7 resulting from the compression of the elastomeric bearing pads at the support. Each girder was 8 instrumented at mid-span with surface mounted resistance-type strain gages placed on the 9 prestressing strands and embedded strain gages placed in the concrete along the section depth. In 10 order to investigate susceptibility of the joint to water seepage before and during loading, three 4 ft 11 (1.22 m) long wood dams placed directly on top of the longitudinal joint were filled with water 12 prior to testing. Detailed information on the instrumentation were provided by Konrad (7). 13

Each specimen was subjected to cyclic loading representative of Fatigue I and Fatigue II 14 load combinations (6), then was tested to failure under increasing monotonic load. Fatigue I 15 loading was included in this study to investigate the effects of maximum stress ranges that could 16 result from potential overloads on agricultural routes. The fatigue load combination is based on a 17 single AASHTO design truck and incorporates dynamic load allowance of 15%; however, the live 18 load multiplier in Fatigue II load combination is 0.75 while that in Fatigue I load combination is 19 1.5. Moment envelopes were developed for the design truck using a structural analysis program 20 (8). The analysis was performed for the 33 ft. (10.06 m) wide hypothetical bridge described earlier 21 and using the AASHTO’s load distribution factors (6) for an interior girder. The load for the 22 fatigue testing was determined by matching the maximum moment resulting from the fatigue live 23 load and the point load used during testing. Thus, the fatigue loading consisted of a cyclic point 24 load having amplitudes of 21 kips (96.7 kN) and 42 kips (187.3 kN) for Fatigue II and Fatigue I 25 load combinations, respectively. The loading was applied at a rate of one cycle per second for 26 Fatigue II and 0.75 cycle per second for Fatigue I. Based upon expected average daily truck 27 traffic, the number of load cycles corresponding to 75 year of service was determined to be 28 411,000. A strength test was performed for each specimen following the completion of the 29 respective fatigue loading. The test setup for the strength tests was identical to that used for the 30 fatigue tests. The strength test loading was quasi-static with displacement increments ranging 31 between 0.02 in (0.51 mm) during the initial elastic response and 0.05 in. (1.27 mm) after 32 significant deflection had occurred. It should be noted that damage to the specimens during 33 fatigue loading were repaired prior to the start of the subsequent testing regiment. 34 35

A B

Wehbe, Konrad, Breyfogle 8 EXPERIMENTAL RESULTS 1 During the fatigue testing, periodic quasi static load tests were performed to assess the effect of 2 fatigue loading on stiffness, relative deflection at the joint, and joint rotation. The static load was 3 equal in value to the cyclic load amplitude. The measured specimen stiffness was determined as 4 the static load divided by the mid-span deflection at the center of girder A. 5 6 Effect of Fatigue Loading on Stiffness and Joint Performance 7 8 “Conventional” Specimen 9 Fatigue I loading was performed first. The first sign of joint deterioration appeared at 15,000 10 cycles when water from the water dam closest to mid-span started to seep through the longitudinal 11 joint. At 31,500 cycles, the welded connection located at 2.5 ft (0.76 m) south of mid-span failed 12 when the weld connecting the welded plate and the angles cracked longitudinally along the length 13 of the connection. The same type of failure occurred at the connection 2.5 ft (0.76 m) north of 14 mid-span at 37,500 cycles. The welded connection at 7.5 ft (2.29 m) south of mid-span failed at 15 44,000 cycles. At 56,000 cycles, connection failure occurred at 7.5 ft (2.29 m) north of mid-span. 16 The test was stopped at 56,000 cycles because the specimen was no longer capable of transferring 17 shear adequately across the longitudinal joint. 18

The longitudinal joint of the “Conventional” specimen was repaired to its original 19 condition before the start of the next fatigue test. Prior to applying the 21 kip (96.7 kN) fatigue 20 loading, the stiffness of the specimen was measured to verify that the stiffness was not affected by 21 the previous fatigue loading and the longitudinal joint repair. At 19,500 cycles water began to seep 22 through the longitudinal joint close to mid-span. Similar to the first fatigue test, the leak spread 23 farther away from mid-span with increasing number of load cycles. At 62,000 cycles, the first 24 welded connection failed at 2.5 ft (0.76 m) north of mid-span when the weld between the angle and 25 the connection plate cracked. The connections at 2.5 ft (0.76 m) south of mid-span and 7.5 ft (2.29 26 m) north of mid-span broke at 67,000 (12.2 service years) and 80,000 cycles (14.6 service years), 27 respectively. The test was stopped at 80,000 cycles as the specimen was no longer able to 28 adequately transfer shear between the two girders. 29

The measured stiffness of the specimen is plotted against the number of load cycles in 30 Figure 6. For the 21 kip (96.7) fatigue load (Fatigue II), the stiffness deterioration rate was 0.13 31 kip/in (0.0228 kN/mm) per 1000 load cycles. For the 42 kip (187.3 kN) fatigue load (Fatigue I), 32 the stiffness deterioration rate was 0.41 kip/in (0.0720 kN/mm) per 1000 load cycle, more than 3 33 times that for the 21 kip fatigue load. 34 35

FIGURE 6 Measured stiffness for the “Conventional” specimen. 36

Wehbe, Konrad, Breyfogle 9 1

Figure 7 shows the measured relative deflection and joint rotation. For the 21 kip (93.7 2 kN) fatigue load, the relative deflection and the joint rotation increased by 0.05 in. (1.3 mm) and 3 0.13 degrees, repectively when the number of load cycles increased from 0 to 80,000. For the 42 4 kip fatigue load, the relative deflection and the joint rotation increased by 0.14 in. (3.6 mm) and 5 0.21 degrees, repectively when the number of load cycles increased from 0 to 55,000. 6 7

FIGURE 7 Measured relative deflections and joint rotation for the “Conventional” 8 specimen. 9 10 “Proposed” Specimen 11 The “Proposed” specimen was first fatigue tested under the 21 kip (93.7 kN) load while all of the 12 rotational restraints (diaphragms) between the stems were in place. This fatigue test was run for 13 500,000 cycles (91.3 service years) with no signs of deterioration to the longitudinal joint or 14 noticeable loss of stiffness. Afterwards, the rotational restraints were removed and an additional 15 200,000 fatigue cycles (36.5 service years) were performed under the 21 kip (93.7 kN) load. Even 16 without rotational restraints, there were no signs of joint deterioration or loss of stiffness during the 17 additional 200,000 cycles. It was concluded that the restraints were redundant and did not provide 18 tangible benefits to the performance of the modified joint. The “Proposed” specimen was then 19 subjected to Fatigue I loading (42 kips) with no rotational restraints for an additional 100,000 20 cycles. Again, the specimen did not exhibit any noticeable signs of joint deterioration or loss of 21 stiffness. The measured stiffness is plotted against the number of load cycles for the 21 kip and 42 22 kip fatigue tests in Figure 8. For comparison purposes, the results from the fatigue tests of the 23 “Conventional” specimen are also shown the in figure. Under the 21 kip (93.7 kN) fatigue load 24 (Fatigue II), the stiffness degradation rate was 0.0023 kip/in (0.0004 kN/mm) per 1000 load cycles 25 when the rotational restraints were in place, and 0.0046 kip/in. (0.0008 kN/mm) per 1000 load 26 cycles when the rotational restraints were removed. Although the degradation rate of the 27 restrained case is one-half that of the unrestrained case, both degradation rates are practically 28 negligible. The joint was essentially water tight throughout the fatigue tests. 29 30

Wehbe, Konrad, Breyfogle 10

FIGURE 8 Measured relative deflections and joint rotation for the “Proposed” specimen. 1 2

Figure 9 shows the measured relative deflection and joint rotation. With no rotational 3 restraints, the deck’s relative deflection and the joint rotation were approximately twice the 4 respective values for the restrained cases. However, the magnitudes of the relative deflection and 5 joint rotation for the “Proposed” specimen were practically negligible. Unlike the “Conventional” 6 specimen case, the relative deflection and joint rotation for the “Proposed” specimen were 7 essentially unaffected by the number of load cycles. 8 9

FIGURE 9 Measured relative deflections and joint rotation for the “Proposed” specimen. 10 11 Effect of Joint Detail on Flexural Strength of the Superstructure System 12 13 “Conventional” Specimen 14 The longitudinal joint was repaired to its original condition prior to performing the strength test. 15 The peak load reached during the test was 73.8 kips (329 kN) at a displacement of 1.12 in. (28.4 16 mm) measured at girder A. At a load of 69.4 kips (329 kN) the welded connection 2.5 ft. (0.76 m) 17 north of mid-span failed. The first prestressing strand yielded at the bottom of girder A when the 18 load dropped to 66.1 kips (295 kN) at a displacement of 1.9 in. (48 mm). The welded connection 19 failed when the connecting plate underwent significant double curvature bending as the girders 20 started to separate vertically. This was followed by complete separation between the angle and its 21 embedded headed stud. The peak load of 73.8 kips (329 kN) occurred just before the second 22 connection failed at 2.5 ft. (0.76) south of mid-span. After the second connection failed, the 23 deflection in girder B started to decrease with an increase in the actuator’s head displacement, 24

Wehbe, Konrad, Breyfogle 11 indicating decreased shear transfer across the joint. The test was stopped after the longitudinal 1 joint had completely separated. Figure 10 shows the load-deflection plot and the failed specimen. 2 3

(a) Measured load-deflection (b) “Conventional” specimen at failure

Fig. 10 “Conventional” specimen load-deflection and joint failure. 4 5

The maximum bending moment attained during the test was 722.6 kip-ft (986.6 kN-m). 6 Thus, the conventional joint was capable of developing only 61.9% of the combined theoretical 7 flexural capacity for the two girders before joint failure. 8 9 “Proposed” Specimen 10 The “Proposed” specimen was tested for strength without the rotational restraints. The 11 load-deflection plot for the strength test is shown in Figure 11. The first flexural cracks appeared 12 on the interior stem of girder A (AE) at a load of 28 kips. Although this flexural crack was not 13 noticed during fatigue testing, visual inspection during the static test revealed that the east stem 14 was cracked prior to the test. Girder B initially cracked at the interior stem (BW) at an applied load 15 of 46.3 kips. The maximum load reached during the strength test was 113.1 kips (504.4 kN) at a 16 displacement of 7.2″ (183 mm) measured at the center of girder A. The longitudinal joint remained 17 intact and girder B trailed closely the deflection of girder A during the entire test until flexural 18 failure of the specimen was initiated by crushing of concrete at mid-span in girder A. The failed 19 specimen is shown in Figure 11. 20 21

(a) Measured load-deflection (b) “Proposed” specimen at failure

Fig. 11 “Proposed” specimen load-deflection and flexural failure. 22

Wehbe, Konrad, Breyfogle 12 1

The maximum bending moment attained by the “Proposed” specimen was 1107 kip-ft. 2 Thus, the proposed joint was capable of developing 95.4% of the combined theoretical flexural 3 strength for the two girders, indicating an excellent ability for load transfer to adjacent girders. 4 Under the loading condition considered in this study, the load carrying capacity of the “Proposed” 5 specimen was more than 1.5 times that of the “Conventional” specimen. 6 7 Effect of Joint Detail on Load Distribution to Supports 8 Reactions at the girder stems were measured prior to the application of the fatigue loading. The 9 measurements were done in response to a 21 kip (93.7 kN) quasi static load. The reactions were 10 also determined analytically using finite element analysis software (8) in order to verify the 11 measured values. The analytical modeling details are shown elsewhere (7). Figure 12 shows the 12 experimental and analytical results for each specimen. The results indicate good agreement 13 between the experimental and analytical results. The difference between the experimental and the 14 analytical results ranged between 3% and 13% for the “Conventional” specimen and between 2% 15 and 18% for the “Proposed” specimen. 16 17

Fig. 12 Reaction force distribution to the girder stems. 18 19

For the “Conventional” specimen, the interior stem of the trailing girder (BW) experienced 20 the highest live load reaction while the exterior stem of the trailing girder (BE) experienced a 21 downward reaction that was counteracted by the superstructure’s dead load. The reactions of the 22 trailing girder indicate that the joint acted as a hinge connection which induced an eccentric shear 23 force and an overturning moment on the trailing girder. The measured reaction at the interior stem 24 of the trailing girder (BW) constituted close to 50% of the system’s total reaction, while the 25 reaction at the interior stem of the loaded girder (AE) was only 31% of the total reaction despite the 26 fact that the load was applied almost on top of stem AE. On the other hand, the “Proposed” 27 specimen results indicate that the joint enabled a more uniform distribution of the reaction among 28 the interior stems and better engagement of the exterior stems. When lumped together, the 29 combined reaction at the interior stems (AE and BW) was approximately 60% of the total reaction 30 for the “Proposed” specimens as compared to 80% for the “Conventional” specimen. 31 32 CONCLUSIONS 33 The proposed joint exhibited substantially improved serviceability and strength performance 34 characteristics over the conventional joint. Based on the experimental results, the following 35 conclusions were made. 36

Wehbe, Konrad, Breyfogle 13

The proposed joint mitigated water leakage through the joint. Water leakage through the 1 conventional joint started at 19,500 and 15,000 load cycles for Fatigue II and Fatigue I loads, 2 repectively, whereas the proposed joint remained water tight under 800,000 cycles of 3 combined Fatigue I and Fatigue II loading. 4

The proposed joint almost eliminated stiffness degradation due to fatigue. Under fatigue 5 loading, the conventional joint deteriorated rapidly resulting in significant stiffness 6 degradation, while the proposed joint remained essentially intact and had negligible effect on 7 stiffness. For Fatigue II loading, the stiffness degradation rate of the “Conventional” specimen 8 was 26 times that of the “Proposed” specimen. 9

The proposed joint enhanced the continuity between adjacent girders. The enhanced 10 continuity was manifested by higher flexural strength, increased load transfer across the joint, 11 and more uniform distribution of reaction forces at the girder stems. The flexural capacity of 12 the proposed specimens was more than 1.5 times that of the conventional specimen. 13

While the concept of the proposed joint can potentially be extended to other double tee or 14

voided slab systems, the results presented in this paper are limited to the double tee system 15

investigated in this study. 16

17 ACKNOWLEDGEMENT 18 Funding for this study was provided by the South Dakota Department of Transportation (SDDOT) 19 and the Mountain Plains Consortium (MPC) University Transportation Center. Cretex Concrete 20 Products is thanked for their contribution throughout the project. The views and findings 21 presented in this paper are those of the authors and do not necessarily represent the views of the 22 funding agencies. 23 24 REFERENCES 25 1. Jones, H. L. Lateral Connections for Double Tee Bridges. Report No. 1856-2. 26

Texas Transportation Institue, College Station, TX. 27 2. Li, L., M. E. Griffey, and R. G. Oesterle. Improved Longitudinal joint details in 28

decked bulb tees for accelerated bridge construction: Concept Development. 29 Journal of Bridge Engineering, Vol 15, No. 3, pp 327–336. 30

3. Li, L., Z. Ma, and R. G. Oesterle. Improved longitudinal joint details in decked 31 bulb tees for accelerated bridge construction: Fatigue evaluation. Journal of 32 Bridge Engineering, Vol 15, No. 5, pp 511–522. 33

4. Li, L., and Z. Ma. Effect of intermediate diaphragms on decked bulb-tee bridge 34 system for accelerated construction. Journal of Bridge Engineering, Vol 15, No. 35 6, pp 715–722. 36

5. Standard Specifications for Roads and Bridges. South Dakota Department of 37 Transportation, Pierre, SD, 2015. 38

6. AASHTO-LRFD Bridge Design Specifications, Fifth Edition. American Association of 39 State Highway and Transportation Officials (AASHTO), Washington, D.C., 2012 40

7. Konrad, M. Precast Bridge Girder Details for Improved Performance. M.S. Thesis, South 41 Dakota State University, Brookings, SD, 2014. 42

8. SAP2000, Version 14. Computers and Structures, Inc. Berkeley, CA, 2012. 43